The Institution of Engineers of Ireland

The Design and Construction of the Boyne Bridge Joe ODonovan BE CEng FIEI MICE MConsEI Managing Director, Roughan & ODonovan Ltd Keith Wilson MA CEng MICE Chief Bridge Engineer, Roughan & ODonovan Ltd Pat Maher BE MEngSc MSc Dip Highway Eng CEng MIEI MICE MIStructE Bridge Manager, National Roads Authority (formerly Chief Bridge Engineer, Roughan & ODonovan Ltd) Paper presented to the Institution of Engineers of Ireland Dublin March 4th 2003 Cork March 11th 2003

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ABSTRACT The Boyne Bridge carries the M1 Northern Motorway over the River Boyne approximately 3 km west of Drogheda. The paper briefly describes the environmental issues which influenced the preliminary design before describing the bridge as finally designed. The bridge has a number of innovative features, such as wind shielding and unpainted enclosed steelwork, and these are described together with the reasons for their inclusion. Other significant design aspects which arise as a direct result of the form of structure are discussed. The procurement strategy for the bridge is described including the development of a supplemental agreement in which the contractor assumed a much greater degree of risk in return for a lump sum payment and the opportunity to develop a deck design which maximised his fabrication and erection expertise. The main differences between the two designs are described. Finally, the key aspects of the pylon construction and the erection procedure for the deck are outlined.

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1. THE NORTHERN MOTORWAY (GORMANSTON TO MONASTERBOICE)

The Boyne Bridge is located on the Gormanston to Monasterboice section of the M1 Northern Motorway. This 22km long section of motorway connects the Balbriggan Bypass to the Dunleer Bypass and is a key element in the major road network of the country. In February 1998, Northconsult, a consortium of consulting engineers was appointed by the client to carry out the detailed design and to supervise the construction stages of the motorway. The consortium consisted of:

M C OSulllivan Ltd Atkins Ltd Roughan & ODonovan Ltd.

The Boyne Bridge was designed by Roughan & ODonovan Ltd as part of the above consortium.

2. THE CLIENT The proposed Northern Motorway involved a number of Local Authorities Meath County Council, Drogheda Corporation and Louth County Council. The latter two authorities, however, under a section 59 agreement of the Local Government Act (1995) confirmed that Meath County Council would exercise all their functions in relation to the motorway and its associated link roads.

3. SITE OF THE BRIDGE The Boyne Bridge is located about 3km west of Drogheda, see Figure 1. The bridge crosses the river approximately at right angles in a north-south direction, see Figure 2. The River Boyne is approximately 150m wide and consists of two channels separated by Yellow Island. The river is tidal at the crossing. At the bridge, the river valley has an asymmetrical cross-section with a steep escarpment to the south of the river and a gently rising slope from the river edge going northwards.

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Figure 1 Bridge Location

Figure 2 Bridge Site

4. MOTORWAY AT THE BRIDGE The motorway is about 20m above the normal water level of the river and the distance between the points where it departs from ground level on either side of the valley is approximately 400m. The bridge cross-section will accommodate two carriageways, each with a road paved width of 11.5m in the short term and 12.75m in the long term, see Figure 3. The short term width of 11.5m comprises a 3.0m

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hard shoulder, 2 x 3.75m wide lanes and a 1.0m wide shoulder on the median side. In the long term, provision is made for an additional 3.75m wide lane by reducing the width of the median, reducing the hard shoulder width to 0.5m and adjusting the lane positions.

Figure 3 Motorway Cross-section at Bridge

5. PRELIMINARY DESIGN OF THE BOYNE BRIDGE

The preliminary design of the Boyne Bridge was completed in June 1994 and in April 1995 the Environmental Impact Statement (EIS) was published. The EIS for the Boyne Bridge was separate from that done in respect of the motorway and was the first formal bridge EIS to be carried out under the Roads Act 1993 and Part 5 of Roads Regulations 1994. A paper on the preliminary design of the bridge was presented to the North-East Region IEI in November 19961. That paper detailed the environmental issues and the constraints which influenced the bridge design ultimately selected and these may be summarised as follows: (a) Historical and Heritage The Boyne Valley is one of the most historically significant sites in Ireland. The ancient monuments of Newgrange, Knowth and Dowth lie within 3-5km of the crossing. The three main Williamite crossings of the river at the Battle of the Boyne in 1690 took place adjacent to the bridge location. (b) Flora and Fauna The EIS concluded that the site for the proposed bridge represents the most ecologically important area in the proposed motorway. Eight

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distinct habitats were identified within the environs of the river and all show a diversity of flora and fauna species. (c) Amenity Walkers and anglers make considerable use of the southern bank of the river along the Oldbridge Road. (d) Physical Within the river channels, on Yellow Island and in the tidal reed beds to the north of the river, the presence of soft alluvial conditions, mudflats and fluctuating water levels would make the construction of bridge substructures very difficult irrespective of environmental considerations. (e) Aesthetic Given the outstanding natural beauty of the Boyne Valley, the scale and height of the bridge at the crossing, the aesthetic merit of any design was regarded as probably the single most important factor to be considered. Accordingly the design team took the view that the bridge design should as far as possible achieve the following aims:

be compatible with the river valley, make a positive visual statement, cause minimal interference with flora and fauna, cause minimal intrusion onto the Battle of the Boyne Site, cause minimal ( even if temporary) loss of amenity during

construction, and be the design which best meets the design constraints and not

necessarily the cheapest design.

6. GENERAL DESCRIPTION OF BRIDGE Span Arrangement The elevation of the bridge is shown in Figure 4. The deck spans, centre to centre of bearings, are as follows:

Back span 42.5m Main (cable stayed) span 170.0m

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Side Span 1 45.0m Side Span 2 40.0m Side Span 3 30.0m Side Span 4 25.0m Total Length 352.5m

Figure 4 Elevation of Bridge

Pylon The pylon is an inverted Y in elevation, extends some 93m above the top of the pile cap and is constructed in reinforced concrete. The legs and the pylon head are hollow being generally rectangular in section and tapering in both front and side elevations. The pylon legs vary in section from 5.0m x 4.8m at the base to 4.2m x 3.865m at the underside of the pylon head. Back stays and fore stays are stressed and anchored from the inside of the pylon head. Full access for both personnel and equipment is provided to the pylon head through the hollows in the pylon legs and head. There is no access from the pylon into the deck void. The legs of the pylon are connected with a concrete crossbeam at underside of deck level. The cross beam is rectangular in section 4.0m wide x 3.9m deep, is hollow inside and supports the deck at two points transversely through mechanical bearings. The 7m x 7m x 3.5m deep concrete plinths at the bottom of each leg of the pylon are connected across with a 4m wide x 2.5m deep concrete tie beam. Each plinth is supported by a 17.5m x 16m x 2.5m deep reinforced concrete pile cap while each pile cap sits on 16 no. 1.5m diameter concrete bored piles.

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South Abutment The south abutment provides a gravity anchorage structure for the back stays of the pylon, see Figure 5. It is a reinforced concrete multi-cellular structure 32.6m x 41.0m x approximately 7.5m high. Many of the cells are fully or partially filled with Class 10/40 concrete. Between the structural concrete and the infill concrete some 17,000 tonnes of

Figure 5 South Abutment

concrete is provided in the abutment. However, permanent access is provided to the interior of the abutment so that the anchorages to the stays can be inspected. The abutment is founded on a spread footing at an average depth of 5m below existing ground level and in service conditions exerts a lower ground pressure than the original insitu condition. Access into the bridge deck is provided from the abutment. Northern Approach Span Substructures The intermediate supports for the northern approach spans consist of pairs of 2m diameter reinforced concrete circular columns. The columns generally decrease in height as the ground gently rises going north, column heights varying from 14m to only 5.5m. Columns are supported on individual reinforced concrete pile caps. The northern abutment is a simple reinforced concrete cantilevered abutment and wing walls which is supported on bored piles. Access into the bridge deck and inspection gallery for bearings and the movement

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joint are provided at this abutment. All piles for the northern approach span substructures are 750mm diameter. Bridge Deck The cross-section of the deck as designed, shown in Figure 6, is basically the cross-section used for all spans of the bridge. The deck structure consists of a 230mm thick reinforced concrete slab acting compositely with a steel space frame giving a structure with high transverse and torsional stiffness. The space frame is mostly constructed from circular hollow sections (CHS) and varies in depth (centre to centre of chords) across the section from 2.7m (on the longitudinal centre-line) to 2.3m.

Figure 6 Bridge Deck

The space frame module is 5.45m transversely and 5.0m longitudinally, the latter dimension matching the 10m stay cable spacing. The deck possesses high torsional stiffness resulting in good aeroelastic stability. The majority of nodes where members converge can be fabricated by profiling the ends of the members and welding them together. At some positions, however, cast steel nodes are necessary because of the effect of node geometry on the design stress levels. The manufacture of cast steel nodes is a specialist activity and the design of the nodes would be carried out by the specialist supplier of the castings. All of the structural steel is enveloped by a pultruded glass reinforced plastic (grp) enclosure so that conditions close to indoor prevail within the enclosure. It is expected that, in these conditions, the steel will require very low or possibly no maintenance in the future. The enclosure provides a ready made access platform for the future inspection of the steelwork, deck soffit and movement joints.

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Stay Cables The main span of the deck is supported by an inclined plane of stay cables at each side of the deck. The fourteen stays on each side fan out in a semi-harp arrangement from the top of the pylon and pick up the deck at 10m centres with a 22.5m gap between the pylon and the first stay and a further 17.5m gap between the last stay and the pier support at the end of the main span. The semi-harp arrangement of stays provides reasonable space for stressing purposes whilst maintaining the advantage of steeper inner stays. The number of back span cables are the same as those of the main span cables and their lines of action intercept on the centre line of the pylon. The back stay cables are arranged in two parallel groups in a harp arrangement and are anchored at each side of the abutment structure. The inclined geometry of the cables together with the vertical line of anchorages in the pylon result in a visually interesting warped plane of the cables on each side of the deck. This feature, in conjunction with the portal character of the pylon produces a particularly striking appearance to motorists approaching the bridge from north to south. The design envisaged the use of either parallel strand or parallel wire stay cables. Each parallel strand stay was to consist of:

(a) a group of 15.7mm diameter seven-wire galvanised strands, each strand enclosed within a wax filled high density polyethylene (hdpe) coating, and

(b) a hdpe outer protective sheath, resistant to weathering and ultraviolet light radiation, surrounding the grouped strands.

Each parallel wire stay cable was to consist of:

(a) a group of 7mm diameter galvanised prestressing wires, (b) a petroleum wax corrosion inhibiting medium surrounding

the wires and filling the space inside the outer protective sheath, and

(c) a hdpe outer protective sheath, resistant to weathering and ultraviolet light radiation, surrounding the grouped wires.

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7. DESIGN OF THE BRIDGE Wind Considerations The prevailing winds at the site of the Boyne Bridge blow from a south westerly or westerly direction, i.e. along the Boyne Valley, which has a funnelling effect on the wind speeds. The road deck of the bridge is 20 m above the river level. It is therefore probable that the wind conditions on the bridge will be worse than elsewhere on the motorway and could cause problems for drivers of wind susceptible vehicles. Experience on the West Link Bridge in Dublin and the Second Severn Crossing in Britain has shown that the provision of porous windshields along the edges of the bridge deck can significantly reduce the effects of high winds on traffic using the bridge without having too much of an adverse effect on the aerodynamic performance of the bridge. The issues involved are complex and are succinctly described by Wilson2. A specialist study was commissioned from FaberMaunsell Ltd which compared the expected wind regime on the bridge deck with that on the adjacent lengths of motorway and approach roads, and with other exposed crossings in Europe. The periods of restriction likely to be experienced by traffic using the bridge are summarised in Table 1.

Table 1 Likely Restrictions due to Wind (per year)

Limit Ground Level Approach Embankment Bridge Deck

Hrs Days Hrs Days Hrs Days Without Windshields Loss of directional control 124 56 170 63 600 124 WSVs may overturn 5 4 9 6 66 34

With 2m high Windshields Loss of directional control 197 65 WSVs may overturn 8 8

Note: WSV = wind susceptible vehicle The report concluded that:

(a) without windshields or traffic control measures, there is a higher risk of traffic accidents on the bridge compared with an equivalent length of ground level road;

(b) this risk can be significantly reduced by the provision of windshields, to a level approaching that on the approach embankments; and

(c) 2 m high windshields with a 45% porosity should be provided along the bridge.

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The Preliminary Report therefore included a recommendation that wind shielding be provided on the Boyne Bridge because:

(a) the risk of high-sided vehicles overturning would be reduced to a level which would be comparable to that risk along considerable lengths of the Northern Motorway, both existing and proposed. Therefore, the need to divert such vehicles through the existing local road network would not arise;

(b) there will be a reduction in the periods when difficult handling would be experienced by drivers of high-sided vehicles;

(c) there will be no need to rely on observance of warning signs by drivers of high-sided vehicles for implementation of traffic control measures; and

(d) there will be no requirement for enforcement by Garda of traffic control measures.

The layout of the wind shielding is shown in Figure 7. It is 2.1 m high, inclined at an angle of 20 to the vertical to match the inclined cable stays, and has three transparent polycarbonate panels 370 mm wide with 380 mm clear gaps between them giving a porosity of 49%.

Figure 7 Wind Shielding

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The windshields, although porous, present a bluff face to wind blowing over the bridge and reduce the stability of the deck against aerodynamic excitation. It is therefore essential that the aerodynamic response of the deck and windshields is accurately assessed and this is best achieved by carrying out tests on a sectional model in a wind tunnel, which also provides the opportunity to measure the shielding effect of the windshield and the drag forces on the bridge deck. In early 1996 BMT Fluid Mechanics Ltd in Britain was commissioned to undertake a series of tests using a 1:70 scale rigid two-dimensional sectional model in smooth flow conditions, Figure 8. The full-scale

Figure 8 Sectional Model in Wind Tunnel

properties of the deck are given in Table 2. The scope of the tests was to:

(a) determine the critical wind speed for the onset of divergent instability of the deck with windshields at five angles of wind incidence and three levels of structural damping, for two values of flexural/torsional frequency ratio;

(b) determine the vortex shedding response of the deck with windshields;

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(c) determine the static drag force, lift force and moment coefficients;

(d) determine the static pressure distribution around the deck, and (e) determine the windshield efficiency.

Table 2 Full Scale Properties of Deck

Mass 40,890 kg/m Polar mass moment of inertia 5,267,000 kg.m2/m Fundamental bending frequency 0.482 Hz Fundamental torsional frequency 0.699 Hz Overall width 36.0 m Overall depth 3.725 m Centre of twist 3.157 m above soffit

The main aerodynamic mechanisms leading to divergent amplitude oscillations are galloping (or stall flutter) and classical flutter. The former arises on certain shapes of deck cross section due to the variation of drag, lift and pitching moment with the angle of wind incidence or time, while the latter involves coupling between vertical and bending oscillations and depends primarily on the separation of the torsional and bending natural frequencies. This type of motion is potentially catastrophic and must be avoided. Appropriate factors of safety to be applied to the predicted extreme mean hourly wind speed are given in BD49/01 as implemented by the NRA3. For the Boyne Bridge, the critical wind speed for the onset of divergent amplitude oscillations had to be greater than 55 m/s. The results of the wind tunnel tests are summarised in Table 3. The only result which falls below the critical wind speed is for very low damping with the wind inclined at 5 below the horizontal. The site is relatively level transverse to the bridge centre line and there is no reason to expect any severe inclination of the wind. During the studies for the Severn Suspension Bridge, measurements of wind inclination were taken at the site of the Severn Rail Bridge4. These results were reviewed during the studies carried out for the Second Severn Bridge and the following conservative generalisation proposed for the variation of extreme wind speed with angle of incidence:

80% of critical horizontal wind speed (44 m/s for Boyne) at 2.5, and 50% of critical horizontal wind speed (27.5 m/s for Boyne) at 5.

The results for the inclined winds do not exceed these figures and hence the Boyne Bridge with windshields will not be susceptible to divergent amplitude oscillations.

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Table 3 Critical Wind Speeds for Divergent Amplitude Instability

Frequency Ratio = 1.45 Frequency Ratio = 1.38Wind Incidence

(deg)

Torsion Damping (log dec)

Critical Wind Speed (m/s)

Critical Wind Speed (m/s)

-5 0.01 > 80 > 80 -2.5 0.01 > 80 > 80

0 0.01 > 80 > 80 +2.5 0.01 71 66 +2.5 0.04 > 80 > 80 +5 0.01 56 54 +5 0.04 65 62 +5 0.05 68 67

Limited amplitude oscillations may be excited by the periodic cross-wind forces arising from the shedding of vortices alternately from the upper and lower surfaces of the bridge deck. Over one or more limited ranges of wind speed, the frequency of excitation may be close enough to a natural frequency of the structure to cause resonance and consequently cross-wind oscillations at that frequency. The design of the bridge must consider this type of response when there are critical wind speeds below the reference wind speed specified in BD49/01. For the site of the Boyne Bridge, the reference wind speed is 42.4 m/s. The results from the wind tunnel tests are summarised in Table 4 and show that vortex-induced oscillations will only occur when the inclination of the wind is below the horizontal. For the bending mode there were two distinct responses, one at approximately twice the wind speed of the other. The first response is associated with alternate shedding of vortices in the wake of the deck; the second response is associated with the shedding of vortices across the deck due to separation at the leading edge with reattachment further downstream.

Table 4 Vortex Shedding Response (N/Nz =1.45, = 0.04)

Wind Incidence (deg) -3 0 +3 Critical wind speed (m/s) 16.5 Bending

(1st Response) RMS amplitude (mm) stable stable 49

Critical wind speed (m/s) 31 Bending (2nd Response) RMS amplitude (mm) stable stable 76

Critical wind speed (m/s) 28.1 Torsion RMS amplitude (deg) stable stable 0.25

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The tests only modelled the fundamental bending and torsional frequencies, but vortex-induced oscillations can also occur at higher frequencies at similar values of reduced wind speed (V/NzD). Table 5 summarises the response of the structure to aerodynamic excitations using the wind tunnel tests, the natural frequencies calculated during detailed design and BD49/01. Using the research described earlier, the reference wind speed for winds inclined at +3 was estimated to be 0.73 x 42.4 = 31.0 m/s. Hence, the second bending response and torsion will not be excited for the final bridge design. The first bending response will be excited at the first two bending natural frequencies, although the magnitude and number of oscillations are reasonably low, so that fatigue due to vortex-induced oscillations will not be critical. It is worth noting that should the torsional natural frequency fall to 0.7 Hz (the estimated value at the preliminary design stage) the expected dynamic loads could be as high as 80% of the HA loading, although this would occur on less than 100 occasions during the design life of the bridge.

Table 5 Response to Aerodynamic Excitation (wind incidence = +3)

1st Bending 2nd Bending Torsion Reduced critical velocity 9.2 17.3 10.8 RMS amplitude 49 mm 76 mm 0.25 1st Bending Mode, Nz = 0.5 Hz Critical velocity (m/s) 17.1 32.2 Dynamic load (kN/m) 19.8 30.7 No. of cycles in 120 yrs 55,000 2nd Bending Mode, Nz = 0.84 Hz Critical velocity (m/s) 28.8 54.0 Dynamic load (kN/m) 55.8 86.6 No. of cycles in 120 yrs < 100 1st Torsion Mode, N = 0.70 Hz (Preliminary Design) Critical velocity (m/s) 28.1 Dynamic load (kNm/m) 445 No. of cycles in 120 yrs < 100 1st Torsion Mode, N = 0.95 Hz (Final Design) Critical velocity (m/s) 38.2 Dynamic load (kNm/m) 818 No. of cycles in 120 yrs HA Loading (LL 170 m) 120 kN/m 120 kN/m 550 kNm/m

Exceeds reference wind speed Not applicable

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The measured static force coefficients were lower than those given in BD37/885 and were adopted for the detailed design. Measurements were also taken of the static pressure distribution around the deck section in order to obtain local design wind pressures to be used in the design of the enclosure system, Figure 9.

Figure 9 Wind Pressure Distribution

The efficiency of the proposed windshields was investigated by measuring the side force and overturning moment on a model of a furniture van positioned on the deck, Figure 10. Measurements were taken with and without the windshields. The results are summarised in Table 6.

Figure 10 Model Furniture Van

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Table 6 Performance of Windshields

Ratio with wind shields / without wind shields Wind Incidence (deg) Side Force Overturning Moment

-5 0.74 0.89 -2.5 0.59 0.81

0 0.42 0.69 +2.5 0.25 0.43 +5 0.18 0.32

The results are similar to or better than those assumed in the wind shielding study for side force at wind incidences greater than -2.5 and for overturning moment at wind incidences greater than 0. For winds with more negative angles of incidence (i.e. blowing from above the horizontal), the barrier is less efficient. However, as argued above, the ground surface at the site of the bridge is relatively level and there is no reason to expect any inclination of the wind other than that due to its turbulence. Further, reference 4 suggests that although wind inclinations of 5 or more could occur with wind speeds at which loss of directional control is possible, 35 mph (15.6 m/s), these are of very short duration and occur very infrequently. It was therefore concluded that the proposed windshields were appropriate. Foundations The bridge site lies on the northern margins of carboniferous limestone known as the Platin limestone. The Lower Palaeozoic rocks of County Louth lie less than a kilometre to the north. The Platin limestones form a belt extending from the southwest near Duleek, to the north-east through the Platin cement works to Sheephouse and Oldbridge and are folded into a shallow syncline, which plunges towards the southwest. The straight contact between the limestone and the Palaeozoic rocks is indicative of a step and probably faulted boundary. Earlier borehole investigations indicated that the limestones in the area are strongly jointed and fissured, with some of the fissures being filled with a variety of sands, silts and clays. The presence of fissuring and voiding is indicative of the karstic nature of the Platin limestone. The presence of karstic phenomena in the area was well known. Drybridge, to the north east of the site is indicative of a karstic bedrock. During the construction of the Platin cement works, a significant number of fissures and voids were discovered during foundation excavations. Furthermore, during the construction of the Boyne viaduct, considerable foundation problems were encountered in the construction of a river pier because of major variations in rock head level across the foundation, a phenomenon

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again indicative of karstic limestone. Thus all of the evidence available to the design team pointed to the potential for difficult foundation construction arising from the nature of the limestone. The carboniferous limestone is overlain by glacial drift, comprising both gravels and boulder clays, within and in the vicinity of the Boyne Valley. Locally, the river is flanked by low-lying alluvial deposits on the north bank and at Yellow Island. Yellow Island itself is an alluvial spit, which is completely inundated at each high tide, during periods of high flows in the river. It can be inferred that the valley of the Boyne was at one stage deeper and wider than currently, but has been infilled with outwash material from melting glacial ice. Evidence of this is apparent on both banks of the river where gravel pits are exposed. In overall terms the geotechnical characteristics of the site can be summarised as follows:

site underlain by karstic limestone, which was expected to be fissured and voided;

the northern bank of the river and Yellow Island consisting of soft alluvial deposits; and

overburden deposits on the northern and southern approaches to the bridge consisting of a mixture of gravels and boulder clays.

Preliminary site investigations consisting of geophysical investigation and nineteen boreholes were undertaken in late 1993. The cross section of the valley interpreted from the boreholes is shown in Figure 11. On the south side of the river above the escarpment, boreholes indicated a

Figure 11 Geological Longitudinal Section

depth of overburden of 10 to 15 metres. On the riverbank at Oldbridge Road, rock lay close to the surface, at depths of from 1.8m to 4m (0.8 to 2.7m OD). Rock head level fell towards the north reaching depths of 24m (-18m OD) on the edge of the alluvial wetlands. Rock coring confirmed karstic characteristics to some degree in almost all of the boreholes. Empty voids (one up to 5m) were encountered in a number of

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boreholes. In addition, voids filled with clays and sands were encountered. Bands of highly weathered rock, with very poor core recovery, were encountered to considerable depths in the rock. In the wetlands on the northern bank of the river, approximately 5m of soft alluvial and organic material overlay from 10 to 15m of gravels above bedrock. Because of concerns in relation to the presence of fissures and voids within the rock, a foundation arrangement consisting of large diameter bored piles, socketed into rock, was devised. Given the very large loads from the pylon structure, a pile group of sixteen 1500mm diameter piles, each with a working load of 1350 tonnes was used to transmit the load from each pylon leg to the more competent limestone at depth. In the case of the piers supporting the northern approach spans, an arrangement of nine 750mm diameter piles, each with a working load of up to 370 tonnes, was adopted. The northern abutment was also piled to rock using 750mm diameter bored piles. In the case of the southern anchorage abutment, the structure was designed as bearing on the overburden material, on the basis that the bearing pressure under working conditions was little different from the existing pressure at that level in the overburden. Notwithstanding the use of bored piles socketed into rock, and the general improvement of rock quality with depth, the random and unpredictable occurrence of voids and fissures was such that there was always a risk that a void or a zone of very heavily weathered rock could be encountered below the toe of any given pile. Two expedients were adopted in order to deal with this design risk. Firstly, the rock sockets were designed as friction piles only, i.e. no end bearing capacity was assumed in the design of the rock socket. Secondly, a skin friction value of 480 kPa was adopted, a value significantly lower than would normally be used for a rock socket in limestone and reflecting the possible presence of low RQD rock within the socket. In order to confirm appropriate pile design depths and prove that no voids existed in the zone of rock beneath the pile toes, a two stage probing sequence was specified. The first stage or advance probing entailed drilling a series of 54mm diameter probe holes (thirteen in the case of the pylon leg pilecaps and five in the case of the pier pilecaps) to a depth of six times the pile diameter below anticipated toe level so as to confirm the absence of significant voiding below the pile toes. The second stage probing entailed drilling to a depth confirmed by the advance probes at each pile location in order to verify the appropriate toe level.

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Pylon The pylon, see Figure 12, is designed longitudinally as a slender reinforced concrete stayed strut and transversely as a triangular frame. The inverted Y-shape enhances the contribution of the cable stay system to the overall torsional stiffness of the deck.

Figure 12 General Arrangement of Pylon

Transversely the axial loads in the inclined legs are balanced through the prestressed buried tie beam, which limits the horizontal forces transmitted to the piled foundations. The beam is post-tensioned progressively during construction, as the axial load in the pylon legs increases. In order to avoid over-designing the pylon it is imperative that second order (P) effects are properly evaluated6. For the preliminary design,

the first order moment was magnified by a factor of 1

where is the

buckling factor, i.e. the critical buckling load divided by the applied load. For the final design a geometric non-linear analysis was used. In

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both cases it is important to use a flexural stiffness which takes account of creep and the reduction in stiffness of concrete as the ultimate limit state is approached. The cable stays are anchored inside the hollow pylon head. As well as the high local forces, similar to those occurring at post-tensioning anchorages, the horizontal component of the stay forces has to be transmitted across the pylon between the fore and back stays. This produces very high local bending moments and tensile axial loads which have to be carried either by reinforcement in the pylon walls, or by a structural steel assembly, or by a combination of both. The tender design carried these forces using the reinforcement in the pylon walls. The contractor re-detailed the pylon head using a structural steel box assembly which acted compositely with the reinforced concrete, see Figure 13. This had the advantage of not only reducing the density of

Figure 13 Pylon Head Detail

reinforcement in the walls, but also allowed the bearing plates for the stay anchorages to be accurately aligned in the controlled conditions of the fabrication shop. The steel box, or stem liner, was divided into a number of sections vertically so that on site it was only necessary to align each box accurately on top of the previous box for the stay anchorages to be correctly positioned. Such an arrangement requires large forces to be transferred between the concrete walls and the steel box which required careful detailing with 32 and 40 mm diameter reinforcing bars being welded to the steel box and fully anchored into the concrete walls. Personnel access to all levels in the pylon head is provided by access ladders and rest platforms located in the west leg. A pair of rails is provided in the east leg so that a trolley can be used to haul heavy

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equipment up to the base of the pylon head. A lifting beam is located inside at the top of the pylon head so that, for example, jacking equipment can be lifted up should the stays require re-tensioning in the future. The trolley rails are fitted with rungs which allow them to be used to access the aircraft warning light in the east leg and to provide an alternative means of escape to ground level from the base of the pylon head. Ring bolts are provided on the outside of the pylon to allow for abseiling access. Deck The bridge deck is built in to the south abutment, which avoids the need for large thrust bearings to transmit the horizontal component of the stay forces and reduces the sagging moment in the back span. The deck is continuous from the south abutment to the north abutment, where a joint is provided to accommodate movements due to temperature, creep and shrinkage. It is supported vertically at the pylon and at the approach span piers by mechanical pot bearings. One bearing at the pylon, at each intermediate pier and at the north abutment is guided longitudinally and resists transverse loads. The Preliminary Report recommended a tubular steel space frame stiffening girder acting compositely with a reinforced concrete deck slab. This form of construction was chosen in preference to an all-concrete deck as it was some 50% lighter which had a knock on effect of reducing the sizes of the stays and the loads carried by the pylon, piers and foundations. Erection of the composite deck could be carried out using considerably larger segments and in a significantly shorter time than a concrete deck. A space frame produces a particularly rigid structure which increases the torsional stiffness of the deck and therefore improves its stability under aerodynamic excitation. This was considered to be very important as it was intended to provide wind shielding along the edges of the bridge. The modular nature of the space frame lent itself to cantilever erection of the main span and the design was developed allowing for this form of erection. The square configuration of the top chords of the space frame allowed the deck slab to be designed as two-way spanning thereby allowing a thinner deck slab to be used. The other innovative feature of the deck is the use of a pultruded grp enclosure system. This concept had been pioneered by FaberMaunsell on the refurbishment of the Tees Viaduct in Britain in the nineteen eighties. Measurements taken by the Transport Research Laboratory

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(TRL) had shown that the rate of corrosion of untreated steel in such an enclosure was reduced to under 0.02 mm per annum or about 2 mm over the life of the structure. Pultruded grp has excellent long-term durability, which combined with its low weight and high strength makes it very suitable for use as an enclosure envelope for steelwork. However, it is important that the enclosure is almost, but not completely, airtight, as this allows the enclosure to breathe, but not wind and wind-borne pollutants to blow through the deck void. In order to avoid specifying one particular system, a performance specification was prepared incorporating the UK Highways Agency standard BD677. The system was required to carry a safe working load of 2.5 kN/m2, which permits full access for operation and plant for inspection and maintenance of deck steelwork, bearings and deck drainage. The enclosed space is lit throughout and provided with electrical socket outlets for portable power tools. A further advantage of the enclosure is that it gives the deck a smoother, less bluff, cross-section, which partially compensates for the de-stabilising effect of the windshields under wind loading. The bridge is generally designed in accordance with the appropriate parts of BS54008. In addition reference was made to the draft Eurocode for the design of steel structures9 and the CIDECT design guide10. A multi-stay cable-stayed bridge is a highly redundant structure where the permanent vertical load of the deck is balanced by the vertical component of the cable forces. In reality the loads in the stays will vary from those assumed by the designer due to variations in self-weight and adjustments made during construction in order to obtain an acceptable profile. Wilson11 discusses the implications of this. As the effect of these variations depends on the difference between large numbers and because some of them will be self-compensating, it is important that their possible magnitude is not over exaggerated. For the Boyne Bridge the designers adopted the following approach at the ultimate limit state. The pre-load in the stays was defined as the difference between the loads in the stays under permanent loading and those which would be introduced were the permanent loads to be applied as an external load to the final structure. A partial load factor fL of 1.05 or 0.95 was applied to the pre-loads and these were then added to the nominal permanent loads and the partial load factor applicable to the permanent loads was applied to the sum. For the preliminary design the forces in the bridge were determined by linear analysis of a three-dimensional space frame model with the deck

24

modelled as a spine beam. For the final design the model of the deck was expanded to include the individual members of the space frame and the cracked section properties of the deck slab used where appropriate. A linear buckling analysis to determine the buckling ratio, , and the non-linear analysis to determine second order effects were carried out using the LUSAS computer program. The effects of overall temperature changes and differential temperature through the deck were determined in accordance with BD37/885. These were combined with the effects of a 10C difference between the cable stays and the rest of the bridge and between any two opposite faces of the pylon. The gradients within the pylon were taken to be those appropriate for concrete box sections as given in BD37/88. Although the cable stays are located behind P1 bridge parapets there is still a residual risk that one or more stays could be ruptured should an errant vehicle breach the parapet railings, so it is important that the bridge is designed to withstand such an event. Also, there may be a need in the future to replace one or more cables should they become damaged in some way. Therefore, the bridge was designed for the following cable-out scenarios:

(a) Accidental stay rupture The accidental removal of any one stay cable combined with 10% live load under Load Combination 4 at the ultimate limit state only, using partial factors of f3 = 1.0 and m = 1.0. The dynamic effects of rupturing a cable are allowed for by applying forces equal and opposite to twice the load in the ruptured stay to the structure without that stay. (b) Stay replacement The planned removal of any one stay with full live load on the opposite carriageway and a 30 m length of loading of intensity 1.5 kN/m2 adjacent to the stay being removed, applied over the width of one notional lane. This scenario was considered at both the serviceability and ultimate limit states under Load Combinations 1, 2 and 3.

Cable stay bridges are flexible structures and the method of construction is such that local deviations of the deck from its intended vertical line are to be expected. The reinforced concrete edge beams were precast in order not only to obtain a high quality finish, but also to allow them to be finally aligned and fixed in position after the bridge was structurally complete in order to give a good line to the edge of the bridge.

25

Cable Stays The cable stays were designed at the serviceability and ultimate limit states using partial material factors, m, of 2.25 and 1.75 respectively. During erection the maximum stress in the strands forming the stays was allowed to increase to 55% of the characteristic tensile strength of the strands. The fatigue stress ranges in the stays due to traffic load were determined in accordance with BS5400:Part 108 and the rain-flow method used to assess the fatigue life using the design curves from the Post-tensioning Institutes Recommendations12. The overall diameters of the stays are relatively small and they would be susceptible to vortex-induced vibrations should a uniform steady wind blow over a reasonable length of stay. However, because the stays are inclined it is unlikely that the wind speed would be sufficiently uniform over a sufficient length of stay for oscillations to build up. Further, the dynamic response of the bridge and cables depends on the interaction between many parameters and on their actual values in the finished structure. These are impossible to predict precisely at the design stage and therefore it was specified that the detail design of the cable stays should allow dampers to be incorporated at a later stage with little or no modification to the structure, should this prove necessary. A provisional item for the supply and installation of six dampers was included in the tender in order to obtain competitive prices should these subsequently be required. Each stay consists of between 32 and 68 no. 15.7 mm diameter galvanised high-tensile 7-wire strands. The hdpe outer sheathing varies between 180 and 250 mm diameter and was specified with a 2 mm high double helical rib at approximately 600 mm pitch to prevent wind-rain induced vibrations. The critical aspect of cable stay performance is its behaviour under fluctuating loads. In the Boyne Bridge, the cable stay anchorages are effectively fixed to the bridge deck and pylon so that the ends of the cables are subject to flexure under wind, temperature and live loading. Hence, the stays experience variations not only in axial stress, but also in flexural stress. Generally only axial loads are applied to cable stays in standard fatigue tests and therefore any reduction in fatigue performance due to fluctuating flexural loads would not be evident. Hence, three fatigue tests on full-size stays were specified for Boyne Bridge, Table 7, including one in which the anchorage was rotated through 0.5 for one million cycles. The specification stated that historical test data on stays and anchorages of the same construction, assembly and constituent parts

26

and subject to the same stress ranges may be acceptable at the discretion of the Engineer.

Table 7 Stress Ranges for Fatigue Tests (s = nominal tensile strength of wire)

Upper Stress Level

Axial Stress Range (MPa)

Anchorage Rotation Range

Cycles

0.45s 162 2 x106

0.45s 197 0.5 1 x 106

0.45s 197 1 x 106

Recent developments in the design of parallel strand cable stay anchorages has concentrated on attenuating the stresses in each strand due to rotation of the cable within the transition zone immediately behind the anchorage block rather than by the overall behaviour of the deviator, guide pipes and anchorage. This has two consequences. Firstly, useful and relevant feedback on the fatigue performance of the complete stay can be obtained by testing a single strand in a monostrand anchorage, provided the anchorage is of similar construction and longitudinal dimensions to the multi-strand anchorages to be used in the actual bridge. Secondly, such a design allows the deviator to be easily replaced by a friction damper, should damping of the stay prove necessary in practice. The Freyssinet HD2000 stay cable system proposed by the contractor is such a system and an appropriate axial and rotational fatigue test of a monostrand stay and anchorage, similar to the test pioneered by the authors on Taney Bridge13, was developed through discussions between the Engineer and Freyssinet. The test was carried out at the CEBTP testing laboratories near Paris, see Figure 14. There were no wire

Figure 14 Test Rig for Single Strand Stay

27

breakages during the one million fatigue cycles. In addition, Freyssinet submitted historical test data on full size stays of similar construction to those for the Boyne Bridge and, after examination, the Engineer accepted these as meeting the requirements of the specification. Architectural Lighting The asymmetric cable-stayed form of the Boyne Bridge will have a dramatic visual impact and it is believed that sensitive, sympathetic lighting will both reinforce the visual design and help denote this important heritage site within a rural environment. There is an opportunity to highlight and reveal the bridge from a number of standpoints:

as a landmark structure indicating an important site, as a piece of significant architecture, as a balance to the potential dominance of the proposed roadway

lighting, as an interesting feature in the distant views of the area, and as a waypoint on the motorway.

A specialist study was therefore commissioned from lightmatters (formerly Lighting Design Partnership), a British consultancy specialising in environmental and architectural lighting. The proposal which evolved was based on the use of two colours:

the white light of the roadway lighting marking the horizontal deck, and

the lighting of the pylon and cable stays using a deep, subdued colour to bring out the contrast with the roadway.

Such an approach has the following advantages:

it dramatically reduces the potential for light pollution by concentrating the effect on a low level of lighting,

it ensures that the level of necessary lighting associated with the roadway lighting does not dictate the overall level of lighting,

it brings out the form of the bridge in distinguishing between the roadway and the pylon/stays, and

it creates a visual effect that is at once dramatic and interesting. The final choice of colour will be decided following a site test, but the initial proposal envisages using a deep blue or purple/mauve colour. A line of side emitting fibre optics in the architectural colour, located below the edge beam of the deck will produce a fine sharp line to

28

delineate the deck and frame the white roadway lighting. Narrow beam luminaires mounted on the deck and anchorage abutment and directed up the cables will pick out the stays in the deep colour. As the beams converge towards the pylon head they will have the effect of strongly highlighting the bridge structure. Light will be projected up the outside of the pylon from locations at the base and will help to define the pylon shape. The inside faces will be lit in a similar manner providing a dramatic arch for motorists to drive through. Finally, the recess at the top between the pylon legs will be brightly lit to ensure a measure of drama. The luminaires mounted at deck level will be protected from vandalism with stainless steel cages. Where the lights face the oncoming traffic, the cages will be fitted with grp louvres on their ends and traffic faces to supplement the integral cowls and louvres of the luminaires in preventing drivers from being distracted by glare.

8. PILING PROCUREMENT Given the perceived likelihood of difficulties and delays during the piling operation, it was considered prudent to carry out the piling works as a separate operation in advance of the bridge contract. If delays were to occur in the completion of the piling, it were better that these would not also cause a delay to the progress of the main bridgeworks. In June 1998, tenders were sought under the Open Procedure. However, tender documents required tenderers to submit specific details in order to demonstrate their technical capacity to carry out the type and size / scale of piling required. Four tenders were received at the end of July 1998 and in mid-September 1998, Ascon Ltd. was awarded the contract in the amount of 2,242,000 (1,765,000) excluding VAT.

9. BRIDGE PROCUREMENT

The construction of larger cable-stayed road bridges had not been carried out previously in Ireland at the time of tender. In these circumstances it was felt that a pre-qualification process should be undertaken in order to establish a list of competent tenderers. The final pre-qualification process, itself would, the design team believed, allow Irish contractors to form alliances with appropriate foreign contractors and specialists. The pre-qualification process was undertaken under the Restricted Procedure of the EU Works Directive 93 / 37 / EEC. To this end, each candidate was required to fill in a Pre-qualification Questionnaire which

29

sought information on a wide variety of topics, such as, company/consortium structure, financial details, technical expertise, manpower, health and safety, sub-contracting and programming. Among the stated pre-qualification criteria for candidates were the following:

(a) minimum annual turnover in civil engineering projects in each of the last three financial years ,

(b) construction of a cable-stayed road bridge within the previous five years,

(c) construction of a steel/concrete composite bridge within the previous five years, and

(d) availability of senior managerial staff with experience in building structures as in (b) and (c) above.

Expressions of Interest were received from eight candidates in mid-September 1998 and at the end of October 1998 the design team recommended that the five best qualified of these be invited to tender for the bridge. At the beginning of December 1998, the Employer invited the five recommended candidates to tender. Tender Documents The tender documents required the tenderer to include the following with his tender:

(a) name(s) of the structural steelwork fabricator, (b) name(s) of the supplier of the enclosure systems, (c) name(s) of the supplier of the steel castings for the bridge deck

steelwork, (d) details of the cable stay system including protective systems and

end terminations, and (e) details of the enclosure system.

Tenders based only on the tender documents were acceptable, i.e. alternative tenders were not allowed. The contract documents were based on the November 1998 Draft of the NRA Manual of Contract Documents for Road Works.14 The time for completion of the works under the contract was 24 months. The estimated cost of the works at the time of the tender was 24.4m (IR19.2m) excluding VAT.

30

Tenders were received at the end of March 1999. Further clarifications were sought from the two lowest tenderers on items such as:

(a) the cable stays, (b) the enclosures, and (c) steel castings

in order to ensure that their proposals were in accordance with the specification. In mid July 1999, Roughan & O'Donovan recommended the acceptance of the lowest tender in the amount of 25.08m (IR19.75m) excluding VAT from SIAC / Cleveland Bridge Consortium. The client was considerably delayed in awarding the tender. The contract award was in fact only made just before Christmas 1999 and the contract itself was signed at the beginning of April 2000 with a date for commencement of the works of 3rd May 2000. Supplemental Agreement Post contract, the Contractor put forward a proposal to alter the layout of the steel deck to suit his own fabrication strengths at his works in Darlington. Put simply, a steel deck consisting of open steel sections in a steel grillage was much more easily fabricated by this contractor than the steel space frame with its hollow sections and cast or fabricated steel nodes. The Employer was only prepared to consider the alternative deck in the context of a Supplemental Agreement in which the final account for the works would be agreed. In mid May 2000 the basis of a Supplemental Agreement was accepted by both parties to the contract. A major reallocation of risk from the Employer to the contractor was an important feature of the Supplemental Agreement. For example, the agreement does not allow additional payment or extension of time for:

full or partial suspensions of the works (e.g. arising from archaeological investigations),

unforeseen physical conditions, instructions which in the Engineers reasonable opinion are

necessary for the satisfactory completion of the works or the functioning of the completed works,

legislation enactments and the like, and ambiguities and the like in the contract documents.

31

Basically the only instance where the contractor was entitled to payment was the in case of an enhancement which was defined as a variation other than a variation which in the Engineers reasonable opinion was necessary for the satisfactory completion of the works or the functioning of the completed works.

10. ALTERNATIVE DECK DESIGN The alternative design prepared by the contractor as part of the Supplemental Agreement and checked by the Engineer sought to take advantage of the contractors fabrication and erection strengths whilst maintaining the same external appearance. The deck comprises a ladder beam with the longitudinal girders located on the line of the outer bottom chord of the space frame and cross girders at 3.333 m centres, see Figure 15. The cable stays are connected between pairs of cross girder extensions. The longitudinal girders are 1750 mm deep in the main and back spans, but deepen to 2400 mm at the anchorage abutment and in the approach spans. The bottom flange is up to 1200 mm wide x 100 mm thick with doubler plates at the anchorage abutment, the pylon and Piers 1, 2 and 3. The largest doubler plate, located at Pier 1, is 1100 mm wide by 80 mm thick and extends over a length of 15.4 m.

Figure 15 Alternative Deck

Since there was to be no change to the external appearance of the bridge, the pier locations were not changed. Transversely these were located under the next-to-outer longitudinal chords of the space frame and hence the longitudinal plate girders are out-board of the piers. This induces very large hogging moments in the diaphragm beams at the piers which are carried by twin steel cross girders, 2925 mm deep on the centre-line of the bridge, filled with reinforced concrete and post-tensioned in stages

32

with up to 12 no. tendons, each comprising 27 no. 16 mm diameter 7-wire strands. Each tendon was stressed to a load of 6000 kN. The 230 mm thick reinforced concrete deck slab, which acts compositely with the steel beams, was constructed using precast Omnia planks as permanent participating formwork. The weight of steel in the alternative design is some 30% higher than in the original space frame design. This together with the extra concrete at the approach span piers increases the loads on the foundations by between 10 and 15% at Piers 1 to 3 and up to 26% at Pier 4. Following a review of the trial pile results and an examination of the critical loadcases, it was concluded that the substructures and foundations could carry the additional load without any modification. In addition to the changes to the deck, the contractor also re-detailed the pylon to suit the erection loads for the alternative deck and his preferred method of constructing the pylon. Regions of the pylon where the vertical reinforcement was not required to carry compression were identified, as in these locations alternate vertical bars did not need to be restrained and the transverse link arrangement could be simplified.

11. PILING CONTRACT Works commenced on site in November 1998. Time for completion of the works was five months. The sequence of works involved Ascons probing subcontractor drilling advanced probes. An evaluation of the probes at each pilecap allowed the Engineer to determine the required toe levels of the pile probes at each pile location. Pile probes were taken to a point six times the pile diameter below projected pile toe level. As probe results were received at each pile location, the situation was reviewed and amended toe levels advised to site as required. Such an approach required continuous daily liaison between the Engineers Representative and head office, but was essential in order to take due account of the highly variable conditions which were known to exist on site. Pile toe levels were decided based on an overview of the probe results at each pilecap location. In overall terms, the probing confirmed the results of the site investigations data. Depths to rock head were largely as anticipated, and the rock consisted of alternate bands of competent and heavily fissured / weather rock. At pier locations, P1 and P2, the closest and penultimate piers respectively to the northern bank of the river, very poor rock conditions were encountered. In addition to voiding and heavily fractured rock, interconnection between different probes (as evidenced

33

by air flushing blowing up in adjacent probes and by backfill grout migration into adjacent holes) indicated that voids and cavities extended over a number of metres. Additionally, during a period of very wet weather in January 1999, slight artesian effects were noted with water levels in the boreholes rising to approximately a metre above ground level. Pile construction was undertaken by Dutch piling specialist contractor NGT, a subsidiary of Ascons parent company HBG. A total of 82 no. 750mm diameter rock socketed bored piles were constructed on the northern side of the river, and 32 no. 1500mm diameter rock socketed piles for the pylon foundations on the southern bank of the river. Both the 750mm diameter and the 1500mm diameter piles were undertaken using a Sumito LS 60 tonne crawler crane with a Demag D22 piling hammer, see Figure 16. All piles had a temporary casing extending a minimum of 2m into rock, with an uncased rock socket below.

Figure 16 Installation of Piles

34

Test piling was carried out on two 750mm diameter piles on the north side of the river and one 750mm diameter pile at the pylon. All pile tests were satisfactory. Integrity testing was undertaken on all piles. In the case of the north abutment and piers P3 and P4, pile toe levels were similar to those indicated in the tender documents. For pile groups P1E and P2W, the piles were taken down to levels of -33m and -36m respectively approximately 13m deeper than anticipated in the tender documents, and reflecting the very poor rock encountered. The piling contract was completed in July 1999.

12. CONSTRUCTION General The conforming design envisaged the main span of the bridge being constructed by cantilevering the deck out from the pylon and installing the permanent stays progressively. In his tender, the contractor proposed erecting the main span by launching it from the south abutment and using the pylon to support the deck with two pairs of temporary stays. The length of the temporary stays would be adjusted using strand jacks, while the tension from the hauling cable would match the design stresses in the steelwork. The contractor adopted the same approach for the erection of his alternative deck and refined his design to suit the launching procedure. Pylon The inclined legs of the pylon were constructed in 6 m high pours using the lightweight Alumna falsework system similar to that used on the Charles River Bridge in Boston, USA, see Figure 17. In this system the formwork is supported from a truss, which is constructed between thelegs and is initially supported from the ground. As the legs rise upwards, members of the truss are removed from the bottom and reassembled at the top of the falsework. In this way the falsework climbs with the legs and is supported from the legs. The falsework also serves to keep the legs in their correct lateral positions, obviating the need for jacking the legs apart prior to constructing the pylon head, which would have been necessary if the inclined legs had been constructed in free cantilever.

35

For the pylon head, the contractor adopted 3 m high pours with the steel plates of the stem liners forming the inner faces and the RMD system of climbing formwork the outer faces.

Figure 17 Construction of the Pylon

Work started on cutting down the piles and constructing the pilecaps in September 2000 and the pylon was topped out on 28 March 2002. Deck The erection of the deck is shown schematically in Figure 18. Sections of the longitudinal steel beams of the approach spans were welded together on the ground and lifted into position on the approach span piers. The cross girders were then bolted to the longitudinal girders. The concrete for the composite cross girders at the piers was cast in two stages and the first stage post-tensioning installed. The Omnia planks and reinforcement for the deck slab were placed on the steel framework and the insitu concrete slab poured, starting from the north abutment. The span sections of the deck were poured first followed by the sections over the piers in order to minimise cracking in the slab. As the concrete slab construction advanced, the second and third stage post-tensioning was installed in the pier cross beams. The

36

final stage post-tensioning in the Pier 1 cross beam will be installed after the surfacing has been completed.

Figure 18 Construction Sequence

37

While the approach spans were being constructed, the steelwork for the main and back spans was assembled on stillages behind the anchorage abutment, along the line of the motorway, Figure 19. The outer

Figure 19 Main Span Steelwork

permanent back stay (B14) was installed and used to stabilise the pylon during the launch. The steelwork was transferred to two computer-controlled multi-axle trailers and a skate located at the rear of the anchorage abutment. A temporary post and stays (T1 and T2) were installed to support the leading cantilever until it reached the pylon. A hauling line was anchored behind the north abutment and connected to the leading end of the steelwork. The steelwork was launched by driving the trailers forward and pulling on the hauling line. The trailers also provided braking to control the launch, Figure 18 (b) and (c). When the steelwork had reached the pylon, the temporary post was removed, the skate moved to the back of the abutment and temporary

38

stay T1 connected to the top of the pylon, Figure 18 (d). The launch then continued with the following cycle:

pull, adjust length of temporary stay, survey, and move skate back.

As temporary stay T2 passed the pylon it was also connected to the top of the pylon and the launch continued with two temporary stays, Figure 18 (e). When the steel girders were in the correct longitudinal position, they were lowered to their final levels at the anchorage abutment and pylon and the level adjusted at the closure joint. The steel girders were then welded to anchorage plates in the south abutment and to the end cantilever of the approach spans. The concrete deck was constructed starting at the pylon and working firstly towards the anchorage abutment and then northwards across the main span. As the slab reached the position of the first fore stay (M1), the first back stay (B1) was installed followed by M1, Figure 18 (f). The next 10 m of slab was poured and stays B2 and M2 installed. Construction continued with a typical cycle:

pour 20 m of slab, install two pairs of back stays, and install two pairs of fore stays.

As the construction face reached the temporary stays T2 and T1, they were removed followed by the hauling cable. Finally, the precast edge beams and deck furniture were aligned and fixed in position, the grp enclosure installed and the roadway surfacing laid, Figure 18 (g). Installation of Stays The stays were installed using the Freyssinet iso-tension system of strand-by-strand installation. The hdpe outer sheath is assembled on the deck by welding together standard lengths of sheathing. The first strand is installed and lifted into position with the sheathing. It is then stressed to a predetermined force and attached to the stressing anchorage with a special device incorporating a load cell. The second strand is installed, connected to the anchorages and stressed using a monostrand jack. A second load cell records the force in the second strand, which is stressed until the forces in the two strands are equal. The process is repeated until all the strands are installed in the stay.

39

The initial force in the first strand is determined by the designer taking account of the expected deformation of the structure. The deck is relatively flexible and therefore a small change in fore stay force can result in large vertical movement of the deck. The loads in the stays and the forces and bending moments in the deck can therefore be more precisely controlled by adjusting the length of the stay rather than its load. Each fore stay was initially installed to a load which was consistent with the unstressed length of the cable being 75 mm longer than its theoretical value to ensure that the level of the deck at the stay anchorage would be lower than its required value. The deck was then surveyed, compared with the expected value and the length of cable shortening adjusted if necessary. Each strand was then shortened by this amount. The fore stays were stressed from their upper anchorages located inside the pylon head. Figure 20 shows the expected and actual profile of the deck after structural completion.

The verticality of the pylon was also surveyed after each round of stay installations. Figure 21 shows the theoretical and actual profiles at structural completion. The back stays were stressed from the access galleries within the anchorage abutment. Because the pylon and anchorage abutment present a relatively stiff structure to the stay, it was appropriate to install the back stays to their calculated loads with no subsequent length adjustment.

-200

-100

0

100

200

300

400

500

Rel

ativ

e C

ambe

r (m

m)

Theoretical Actual Corrected for creep + u/s camber

SA Pylon P1

Figure 20 Relative Cambers after Structural Completion

40

A final check of the deck profile will be made after the surfacing has been completed. If required, a final round of stay length adjustments will be made at that stage, but it is expected that this will not be necessary as there has been good agreement between the analysis model and actual behaviour.

13. SUMMARY OF APPROXIMATE QUANTITIES IN BRIDGE

Structural Concrete 19000 cubic meters Reinforcement 2500 tonnes Structural Steel 2200 tonnes 15.7mm diameter strand in stays 333 tonnes (283,000m total length)

14. ACKNOWLEDGEMENTS The authors gratefully acknowledge the permission of Mr. Oliver Perkins, County Engineer, Meath County Council to publish this paper.

Client: Meath County Council Consulting Engineers: Northconsult

Figure 21 Position of Pylon at Structural Completion

0

20

40

60

80

100

-100 -50 0Displacement (mm)

Hei

ght (

m) Theoretical

Survey

Correctedfor creep

SOUTH

41

Consultants for Wind Studies, Wind Shielding and Category 3 check:

FaberMaunsell Ltd

Bridge Contractor: SIAC/Cleveland Bridge J.V Piling Contractor: Ascon Ltd

The authors wish to acknowledge the support and assistance of the following:

Mr. Oliver Perkins County Engineer, Meath County Council

Mr. Charles McCarthy Project Engineer, Meath County Council

Mr. Frank Burke Former County Engineer, Louth County Council

Mr Tony Kearon County Engineer, Louth County Council

Mr Donall OCaoimh and Mr John Iliff

Project Resident Engineers

15 REFERENCES

1 ODonovan, P.J, Preliminary design of proposed Boyne Bridge on Northern Motorway, paper presented to a meeting of North East Region IEI, Nov. 1996

2 Wilson,K.R, Wind resistance design in large scale bridge projects, Long-span bridges and aerodynamics (Ed. Miyata T, et al), Springer, Tokyo,1999.

3 BD49/01 Design rules for aerodynamic effects in bridges, NRA Design Manual for Roads and Bridges, NRA, Dublin, Dec. 2000.

4 Roberts, Sir G, Severn Bridge: design and contract arrangements, Proc. ICE, Vol. 41, Sept. 1968, pp.48.

5 BD37/88 Loads for highway bridges, Design Manual for Roads and Bridges, Highways Agency, London, 1988.

6 ODonovan, P.J, Wilson, K.R and Dempsey, A.T, Second order effects in the design of concrete pylons for cable-stayed bridges, to be published in 2003.

7 BD67/96 Enclosure of bridges, Design Manual for Roads and Bridges, Highways Agency, London, 1996.

8 BS5400:Part 3:1982, Part 4:1990, Part 5:1979 and Part 10:1980, Steel, concrete and composite bridges, British Standards Institution, London 1979-1990.

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9 DD ENV 1993-1-1:1992 Eurocode 3: design of steel structures, Part

1.1 General rules for buildings, British Standards Institution, London 1992.

10 Wardenier, J et al, Design guide for circular hollow section joints under predominantly static loading (Ed. CIDECT), Verlag TV Rheinland GmbH, Kln, 1991.

11 Wilson, K.R, Safety criteria for major bridges, paper presented to an ICE seminar on Structural Safety, London, July 1999.

12 PTI Committee on Cable-stayed Bridges, Recommendations for stay cable design, testing and installation, Post-tensioning Institute, Phoenix, Aug, 1993.

13 ODonovan, P.J, Wilson, K.R, and Dempsey, A.T, The design and construction of Taney Bridge, Dundrum, IStructE (Republic of Ireland Branch), Dublin, Jan, 2003.

14 Manual of contract documents for roadworks, National Roads Authority, Dublin 1998 (draft)