deflection assessment of an frp-reinforced concrete...

Stone, D., A. Prota, and A. Nanni, "Deflection Assessment of an FRP-Reinforced Concrete Bridge, " American Concrete Institute Convention, Spring 2002, Detroit, MI, April 21-26, 2002.

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Deflection Assessment of an FRP-Reinforced Concrete Bridge

By Danielle K. Stone, Andrea Prota, and Antonio Nanni Synopsis: Serviceability of FRP-reinforced concrete structures remains a highly relevant issue as more structures are constructed using the technology. With the recent publication of the ACI-440H document “Guide for the Design and Construction of Concrete Reinforced with FRP Bars,” the need to examine serviceability-related issues and validate the accuracy of these design guidelines is heightened. A short-span concrete slab bridge was constructed in St. James, Missouri, using precast panels reinforced with FRP bars. The bridge was designed to meet AASHTO load and deflection requirements using the “Guide for the Design and Construction of Concrete Reinforced with FRP Bars.” Carbon FRP, as tensile reinforcement, and glass FRP, as shear reinforcement, were utilized.

Laboratory testing of one bridge panel that is identical to those installed in the field was conducted using a 4-point loading configuration. Field testing of the bridge was also conducted to examine its behavior under service load. A loaded dump truck was placed at various locations along the bridge while deflections were measured and recorded. Similar field tests will be conducted annually for the next three years in an effort to monitor the long-term performance of the bridge. The results of the laboratory and field tests are summarized herein; a comparison between the theoretical and measured deflection values is made to illustrate the conservative nature of the prescribed design guidelines. Keywords: bridge structure, in-situ load test, lateral distribution of load, deflection, FRP, reinforced concrete, carbon, glass, design guidelines


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ACI Member Danielle K. Stone is a doctoral candidate in civil engineering at the University of Missouri-Rolla, Rolla, Missouri. She has been a member of ACI since 1999. Andrea Prota is a master’s candidate in civil engineering at the University of Missouri-Rolla, Rolla, Missouri. In April 2001, he was awarded with the Outstanding Graduate Paper Award by the Missouri Chapter of the ACI. As part of the award, he received a one-year membership to the Missouri Chapter of the ACI. ACI Fellow Antonio Nanni is V&M Jones Professor of civil engineering at the University of Missouri-Rolla, Rolla, Missouri. He is a member of ACI Committees 437 Strength Evaluation of Existing Concrete Structures, 440 Fiber Reinforced Polymer Reinforcement, 530 Masonry Standards Joint Committee, and 544 Fiber Reinforced Concrete. His research interests include the performance of concrete-based structures.

INTRODUCTION

An aging and deteriorating infrastructure has prompted government leaders and engineers to consider new construction technologies to enhance life span and strength of bridge structures. Advanced composites made of fibers embedded in a polymeric resin, also known as fiber-reinforced polymer (FRP) materials, have recently emerged as a viable and practical construction material. The acceptance of FRP materials into mainstream construction, however, has been hindered by various barriers including the lack of design guidelines and codes. The recent publication of the ACI-440H document “Guide for the Design and Construction of Concrete Reinforced with FRP Bars” is a major step toward the establishment of accepted design protocols.

One issue that requires additional attention with the design of FRP-

reinforced concrete is serviceability. Due to the decreased stiffness of FRP reinforcing bars, as compared to steel reinforcing bars, crack width and deflection can become the controlling factors of the design. This paper investigates the serviceability-related issues involved in the design of FRP-reinforced concrete (FRP-RC) structures by examining the performance of FRP-RC bridge panels in-situ and in the laboratory.

The overall project scope includes the procurement of the design, manufacturing, and installation of FRP composite bridge panels for three bridges and FRP-reinforced concrete panels for one bridge. All four bridges are located in a residential area of the small community of St. James, Missouri. Of the three FRP composite bridges, one is a full FRP composite bridge, while the other two are FRP composite bridge decks supported by steel stringers. One of the stringer-


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supported bridge decks consists of longitudinal panels, the other of lateral panels. The full FRP composite bridge illustrates the use of this technology for new construction. The FRP composite bridge decks illustrate the possibility of using this technology for bridge deck replacement. A review of the installation and testing conducted on the FRP panel bridges can be referenced in Stone et. al. (1). The FRP-reinforced concrete bridge consists of longitudinal panels reinforced with FRP bars and illustrates the potential for new bridge construction. Both types of construction consist of bridge deck panels that are pre-manufactured, transported to the site, and assembled on-site.

This paper focuses primarily on the deflection of FRP-RC bridge panels as measured in laboratory testing of individual panels and in-situ testing of the constructed bridge. Comparisons will be drawn between the theoretical deflection values as predicted by the ACI-440H design guidelines and the experimental values and an investigation of the lateral load distribution between panels will be conducted. A brief description of the installation of the bridge is also provided for background.

DESIGN

Utilizing the American Concrete Institute’s document “Guide for the Design and Construction of Concrete Reinforced with FRP Bars,” (2) the bridge was designed to meet the load and deflection requirements of the American Association of State Highway and Transportation Officials. It was designed to carry a standard HS20-44 (approximately 180-kN) truck loading within the span length divided by 800 deflection requirement. Strength reduction factors, φ, of 0.7 and 0.85 were used for the flexure and shear design, respectively.

For the flexural design, the procedure is very similar to that used in the

case of steel reinforcement once the appropriate modes of failure are recognized. The two possible failure modes are (a) rupture of the FRP reinforcement prior to crushing of the concrete and (b) crushing of the concrete prior to rupture of the FRP reinforcement. According to ACI-440H both modes of failure are acceptable; due to the linear-elastic behavior of FRP materials up to failure, the concrete crushing failure is marginally preferred because a small degree of plasticity is exhibited prior to failure. Compared to the flexural design strength reduction factor used in steel-reinforced concrete design (i.e., 0.9), the factor used for FRP-reinforced concrete is 0.7 to account for the decreased ductility of the section. In this specific situation, the section is over-reinforced to assure concrete crushing prior to rupture of the FRP bars.

For the shear design, the separate contributions to the shear capacity of the

concrete, Vc, and the reinforcement, Vf, are still considered. However, the ratio f f s sE Eρ ρ is used as a multiplier of the concrete shear strength contribution to

account for the reduced stiffness of the FRP bars compared to steel reinforcing


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bars; the reduced stiffness will lead to larger shear cracks thereby decreasing the contribution of the concrete to the shear capacity. It should be noted that ρf is the reinforcement ratio of the FRP-reinforced section, Ef is the longitudinal modulus of elasticity of the FRP reinforcement, ρs is the reinforcement ratio of a steel-reinforced section of equal capacity, and Es is the longitudinal modulus of elasticity of the steel reinforcement. It should be underlined that the term “equal capacity” refers to the design moment capacity, φMn.

The panel reinforcement consists of the aforementioned commercially available carbon FRP (CFRP) and glass FRP (GFRP) reinforcing bars with the following properties. For the CFRP bars, a guaranteed design tensile strength of 1862 MPa (270 ksi) and a tensile elastic modulus of 104.7 GPa (15.2 Msi) were given by the manufacturer. For the GFRP bars these values were 725 MPa (105 ksi) and 42 GPa (6 Msi), respectively. A compressive strength of 55.1 MPa (8000 psi) was used for the concrete. Verification of the FRP material properties was conducted and is outlined in the “Laboratory Testing” section.

Based on the material properties and design parameters, the longitudinal reinforcement consists of 12 bundles of three 9.5-mm (3/8-in) CFRP bars, while for the shear stirrups 9.5-mm (3/8-in) GFRP bars are utilized. Although their contribution to the flexural capacity of the member is not considered, 12.7-mm (1/2-in) GFRP bars are utilized in the top side of the cage. Transverse reinforcement consisting of 12.7-mm (1/2-in) GFRP bars at 1.2 m (4 ft) is also provided in the panels. Due to the fact that thermoset resins are used in the manufacturing of the FRP bars, the FRP manufacturer conducted all necessary bending of the reinforcing bars prior to their shipment to the concrete precaster. Figure 1 illustrates the layout of the FRP reinforcement in each panel.

INSTALLATION SUMMARY

The Walters Street Bridge consists of nine precast concrete panels, each with a depth of 0.30 m (1 ft) and a width of 0.86 m (2.83 ft). Overall, the bridge measures 7.32 m (24 ft) in span and 7.77 m (25.5 ft) in width and has a skew of approximately 12 degrees. Marshall Industries Composites, Inc. was the manufacturer of the FRP bars utilized in this project and Oden Enterprises, Inc. was responsible for the precasting of the bridge panels and the installation of the bridge.

Installation proceeded based on the following outline of tasks: • Install precast concrete panels – place panels onto the abutments (see

Figure 2); connect panels; secure the panels to the abutments; fill the panel joints.

• Install the bridge guardrails – attach the guardrail posts to the panels; attach guardrail to posts.


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Tasks unique to the use of precast concrete panels that warrant further attention are the connection of the panels to one another, the connection of the panels to the abutments, and the connection of the guardrail posts to the panels.

The connection of the panels together was accomplished through the use of steel angles, which were embedded in the panel edges and then welded together once the panels were in place on the abutments. See Figure 3 for a detail of the angles just before welding. It should be noted that this is the only steel detail left in the panels and work is currently in progress to develop a non-metallic means of connection. For the connection of the panels to the abutments, a void was formed approximately 0.15 m (6 in) from each end of each panel, through which a hole was drilled into the abutment to receive an anchor bolt. The anchor bolts were secured into the abutments with a two-part epoxy and a nut was tightened down to secure the panels. Once the panels were connected together and to the abutment, non-shrink grout was used to fill the joints and to cover the anchor bolts. The connection of the guardrails to the panels was accomplished using the same type of steel angles embedded at the panel joints. See Figure 4 for a detail of the guardrails welded to the panels. Once the guardrail posts were welded to the panels, the guardrails were bolted to the posts completing the installation of the bridge.

LABORATORY TESTING

Characterization of the FRP bars was conducted in the laboratory to verify the tensile properties of the FRP bars provided by the manufacturer. Two bars of each type were tested, with the results detailed in Table 1. To avoid failure of the specimen at the grips due to the relatively low transverse strength of the FRP materials, the ends of the specimens were encased in steel pipe using an expansive grout. A gripping length of 38.1 cm (15 in) was used based on work conducted by Micelli et al. (3). Furthermore, an overall specimen length of 40db plus two times the gripping length, where db is the diameter of the bar, was used based on provisional specifications for FRP bars testing that are under review by ACI committee 440K.

For both the carbon and glass FRP bars, the tensile modulus and tensile

strength measured in the laboratory were within approximately 10 percent of the manufacturer’s recommended values. Table 2 summarizes the results of the testing. For the CFRP bars, the tensile modulus and tensile strength were approximately 10 percent higher than the values reported by the manufacturer. For the GFRP bars, a similar trend was observed with a higher tensile strength. However, the measured modulus values were approximately 12 percent lower than those reported by the manufacturer.

Testing of the FRP-RC panel was conducted under four-point bending. Figure 5 illustrates the test setup. The 7.32-m (24–ft) specimen was tested over a clear span of 6.4 m (21 ft) with the equal loads applied approximately 2.74 m (9


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ft) from each support, leaving a constant-moment region 0.91 m (3 ft) in length. The section was over-reinforced indicating that failure would occur by crushing of the concrete prior to FRP rupture. The beam was tested to failure through the application of quasi-static load cycles. Several load cycles were conducted; each load cycle proceeded up to the desired load followed by unloading to approximately 22.24 kN (5 kips). During the final load cycle the specimen was taken to failure.

In addition to the linear variable differential transformer (LVDT) transducers and electrical resistance strain gages installed on the exterior of the specimen in the laboratory, electrical resistance strain gages were installed on the reinforcing bars prior to casting of the panel. Strain gages were located on several of the tension longitudinal reinforcement and compression longitudinal reinforcement bars at mid-span and on the shear stirrups at 0.51 m (20 in) and 0.76 m (30 in) from one end of the panel. The results of the testing are summarized herein in terms of the load-deflection and moment-curvature diagrams and the observed failure modes. The load-deflection diagram for the panel is illustrated in Figure 6. The experimental load-deflection relationship is plotted versus two theoretical load-deflection relationships. The curve labeled “bi-linear” was obtained via double integration of a bi-linear approximation of the moment-curvature relationship and the curve labeled “ACI” was obtained based on the experimental load and the deflection calculation guidelines provided by ACI-440H. It should be noted that the difference between the initial slope of the experimental curves and the initial slope of the theoretical curves is due to the presence of a crack near mid-span that must have occurred during shipment. The slope of the bi-linear curve and the experimental curve after cracking are nearly identical, however the ACI-440H theoretical curve exhibits a much lower stiffness than the other two curves. A comparison of the experimental deflection, a maximum of approximately 107 mm (4.2 in), and the ACI-440H theoretical deflection, a maximum of approximately 179 mm (6.9 in), indicates that the experimental deflection is approximately 60 percent of the theoretical deflection as predicted by the ACI-440H guidelines.

Figure 7 illustrates the moment-curvature diagram for the panel. The experimental moment-curvature relationship and the bi-linear approximation of the theoretical moment-curvature relationship are both plotted. The same trends are exhibited in the moment-curvature curves that were noted with the load-deflection curves. Namely, the crack in the section accounts for the difference in the curves at the beginning of the test, however once completely cracked the stiffness of the experimental and theoretical curves is nearly identical. The failure mode exhibited by the panel was shear failure at a load of approximately 209.1 kN (47 kips). The shear cracking (highlighted by the lines drawn on the figure) and the crushing of the concrete that occurred at failure can be seen in Figure 8.


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Figures 9 through 11 illustrate the load versus strain in the tensile reinforcement, the strain in the compression reinforcement, and the compressive strain in the concrete, respectively. Three electrical resistance strain gages at different lateral positions were located at each depth, however one of the gages on the compression reinforcement failed to work properly. The trends exhibited in these plots are further examined by considering several normalized parameters: strain in the tensile reinforcement divided by ultimate strain at rupture, compressive strain in the concrete, and depth of the neutral axis divided by entire section depth. Strain in the tensile reinforcement increases bi-linearly until panel failure at a strain of approximately 0.005. Rupture of the CFRP reinforcement would occur at an ultimate strain of approximately 0.015. Failure of the panel occurred when the strain in the CFRP reinforcement was approximately 33 percent of the ultimate strain at rupture indicating a stress in the tensile reinforcement of roughly 552 MPa (80 ksi). The maximum compressive strain in the concrete at failure was approximately 0.002, or nearly 66 percent of the theoretical maximum concrete compressive strain, which is generally taken to be 0.003. The location of the neutral axis was determined by examination of the strain in the concrete and FRP reinforcement in the panel. The location of the neutral axis of the section at failure, expressed as a percentage of the section depth, was approximately 0.223. The theoretical location at failure of the panel was 0.213, which is very close to the experimental failure ratio.

Further testing of the panels will be conducted in order to determine the shear capacity of the panels. The portions of the panels near the support that are relatively uncracked will be re-tested to induce shear failure. Further verification of the compressive strength of the concrete will also be conducted by coring the panel following the testing to determine the shear capacity.

FIELD TESTING

Shortly after installation of the bridge, the behavior of the bridge under load was examined. A picture of the bridge during the load test is shown in Figure 12. Instrumentation utilized during the testing included direct current variable transformer (DCVT) transducers, which were installed underneath the bridge to monitor deflection of the bridge panels. The location of the DCVT transducers is illustrated in Figure 13, with the symbol denoting each individual instrument. Six DCVT transducers were located at mid-span and six were located near the supports.

Loading of the bridge was accomplished with a loaded tandem-axle dump truck placed at various locations on the bridge. Table 3 outlines the truck’s axle spacing. The total weight of the truck was 212.98 kN (47,880 lb) with 66.19 kN (14,880 lb), 72.86 kN (16,380 lb), and 73.93 kN (16,620 lb), on each of the three axles from the front to the rear of the truck, respectively. Several passes of the


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truck were made, each at a different transverse position on the bridge. Figure 14 illustrates the lateral location of the first four truck passes. Additional passes were made at 32 kph (20 mph) at the same location as Pass 4 as were three passes that were symmetric to Passes 1 through 3. Assuming that the bridge behaved symmetrically, the measurements from the symmetric load passes were used to complete the deflected shapes for Passes 1 through 3. During each pass the truck was stopped at five longitudinal locations. Table 4 details the location of the truck stops. Due to the axle loads and axle spacing of the loading truck, truck location 3 corresponds to the worst-case loading condition.

The results of the load test for Passes 1 through 4 are presented in Figures

15 through 18, respectively. It should be noted that the illustration at the bottom of each figure depicts the layout of each of the nine panels and the lateral location of the tandem axles on the bridge for each pass. Furthermore, the dashed portion of the curve is taken from the abovementioned symmetric pass for each of the passes. For each of the figures, the progression of the deflected shape from the top curve to the bottom curve is consistent with the level of moment induced by each loading position. Stop 1 generates the least moment in the bridge; followed by Stop 1; Stops 2 and 4, which are nearly identical; and Stop 3, which produces the largest bending moment.

A comparison of Figures 15 through 18 illustrates that as the load

progresses from Pass 1 through Pass 4 that the maximum deflection experienced by the bridge decreases due to the fact that a larger number of panels are engaged in sharing the load. A comparison of the maximum deflection during Pass 1 to the maximum deflection during Pass 4 confirms a decrease in deflection of approximately 15 percent.

The impact factor for the live load was examined by conducting a pass in

the same location as Pass 4 at a speed of approximately 32 kph (20 mph) (See Figure 22). The impact factor was computed as the ratio of the deflection obtained at 32 kph (20 mph) to the deflection obtained at Stop 3. The six values, one for each DCVT, were averaged to obtain an impact factor of 0.25. Compared to the computed AASHTO impact factor for this bridge, which is 0.30, the AASHTO guidelines are conservative.

Distribution of load between panels was also examined by comparing the

deflection of the bridge panels to the theoretical load-deflection relationship based on the loading conditions during the load test. Utilizing the deflection obtained in the bridge load test the load experienced by each panel individually was determined. A comparison of these loads quantifies the lateral distribution of load between the panels. It should be noted that the load-deflection relationship for an uncracked section was used. The use of the uncracked stiffness is justified because (a) the load induced during the load test in the panels, a maximum of approximately 12.0 kN (2.7 kips) is approximately 30 percent of the cracking load for the bridge panels and (b) it is unlikely that two fully-loaded trucks would be


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on the bridge at the same time given the surrounding community (i.e., the section should be uncracked).

Figure 19 illustrates the load distribution as a percentage of the total load

on the bridge for Passes 1 through 4. There is a clear progression of the peak load percentage from one side of the bridge toward the center as the load moves from Pass 1 to Pass 4. As was also exhibited in the plots of the deflected shape, it is observed that as the loading truck goes from Pass 1 through Pass 4 the peak load percentage decreases as the number of panels sharing a larger portion of the load increases. Figure 20 illustrates the load distribution as a percent of the total load on the bridge for Pass 4 and the pass at 32 kph (20 mph). Although the total load experienced by the bridge is greater in the case of the 32-kph (20-mph) pass due to impact, the percentage of load carried by each respective panel is very similar. Furthermore, the peak load percentage carried by Panel 4 is identical for the two passes.

Further investigation into the distribution of load between panels, via comparison of the distribution in the bridge to the distribution within a continuous slab, allows the efficiency of the panels to be quantified. Modeling via a commercially-available finite element software package will be conducted. Additional testing in the laboratory will also be conducted whereby three panels, connected in the same manner as they would be in the field, will be tested to evaluate the lateral load distribution. It is envisioned that the deflection obtained in the laboratory will be correlated with a finite element of distinct yet connected panels. Matching the model with the experimental results from the laboratory and then comparing the modeling of connected panels to the modeling of a continuous panel can quantify the degree of efficiency of the load transfer between panels.

Due to the lateral distribution of load between panels the theoretical

deflection is difficult to determine, therefore a direct comparison will not be drawn. It is known however that the bridge panels themselves were designed to meet the AASHTO deflection requirement of span length divided by 800, which in this case corresponds to a deflection of 8.76 mm (0.345 in). The maximum observed deflection during the static load passes was 2.4 mm (0.094 in), yielding a span to deflection ratio of approximately 2940 or approximately one-quarter of the allowable deflections. Even considering the increased deflection experienced during the pass at 32 kph (20 mph), the span to deflection ratio is approximated at 2875. Moreover, Figure 21 illustrates the predicted deflection of the bridge for the design loading condition of one truck in each of the two lanes. The principle of superposition was utilized assuming linear-elastic behavior of the bridge. The maximum deflection in this case is roughly 3.1 mm (0.12 in) for a span to deflection ratio of approximately 2300, or a deflection roughly one-third of the allowable deflection.

In an effort to monitor the long-term performance of the bridge in-situ,

additional field load tests will be conducted annually for two more years. The


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deflection from year-to-year will be compared and any degradation will be quantified. The annual load test will also be combined with an inspection of the visible bridge components for possible wear and degradation.

CONCLUSIONS

Based on the design, installation, and testing of FRP-RC panels conducted the following conclusions can be made:

• The experimental deflection of the FRP-RC panel in the laboratory was approximately 60 percent of the theoretical deflection as predicted by ACI, indicating that the ACI-440H guidelines are very conservative.

• Laboratory testing exhibited good agreement between the experimental and theoretical stiffness values based on moment-curvature predictions.

• The deflection of the bridge panels during the in-situ bridge load test appears to be reasonable with a maximum deflection of approximately 2.5 mm (0.1 in) occurring during the 32-kph (20-mph) pass.

• The maximum measured deflection during the static passes of the load test corresponds to a span-to-deflection ratio of 2940 or approximately one-quarter of the maximum allowable deflection.

• A span-to-deflection ratio of 2300 (or approximately one-third of the maximum allowable deflection) was predicted for the design loading condition of one truck in each of the two lanes of the bridge.

• The load impact factor for the bridge was quantified as 0.25 making the value of 0.30 given by AASHTO a conservative estimate.

• Lateral distribution of the load between panels is occurring and appears to be reasonable. Efforts to quantify the lateral load distribution will continue with future laboratory testing and the initiation of a finite element analysis.

NOTATION Ef longitudinal modulus of elasticity of the longitudinal FRP reinforcement Es longitudinal modulus of elasticity of the steel reinforcement Vc concrete contribution to the shear capacity Vf FRP reinforcement contribution to the shear capacity ρf reinforcement ratio of the FRP-reinforced section ρs reinforcement ratio of a steel-reinforced section of equal capacity, φ strength reduction factor φMn. design moment capacity


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ACKNOWLEDGEMENTS

The authors would like to acknowledge funding and support received from the University Transportation Center at the University of Missouri – Rolla, the Missouri Department of Transportation, the City of St. James, the National Science Foundation, and the Missouri Department of Economic Development.

REFERENCES 1. Stone, D., A. Nanni, and J. Myers. Field and Laboratory Performance of FRP Bridge Panels. In Proceedings of the International Conference on Composites in Construction - CCC 2001, Porto, Portugal, Oct. 10-12, 2001. 2. ACI Committee 440. Guide for the Design and Construction of Concrete Reinforced with FRP Bars. 440.1R-01, American Concrete Institute, Farmington Hills, MI, 2001, pp. 41. 3. Micelli, F. and A. Nanni. Mechanical Properties and Durability of FRP Rods. CIES report 00-22, March 2001.


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FIGURE 1 FRP reinforcement layout

FIGURE 2 Setting the bridge panels onto the abutments


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FIGURE 3 Welded connection of the panels at the joints

FIGURE 4 Installation detail of the guardrails


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FIGURE 5 Test setup for the FRP-RC panel

FIGURE 6 Load-deflection diagram

0

25

50

75

100

125

150

175

200

225

0 20 40 60 80 100 120 140 160 180

Deflection (mm)

Load

(kN

)

experimentalbi-linearACI


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FIGURE 7 Moment-curvature diagram

FIGURE 8 Failure of the FRP-RC panel

0

50000

100000

150000

200000

250000

300000

350000

400000

450000

0 0.00001 0.00002 0.00003 0.00004Curvature (rad/mm)

Mom

ent (

kN m

m)

envelopebi-linear


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FIGURE 9 Load versus strain in the tensile reinforcement

FIGURE 10 Load versus strain in the compression reinforcement

0

25

50

75

100

125

150

175

200

225

0 0.001 0.002 0.003 0.004 0.005 0.006 0.007FRP reinforcement strain

Load

(kN

)

strain gage 6strain gage 7strain gage 8

0

25

50

75

100

125

150

175

200

225

-0.0007-0.0006-0.0005-0.0004-0.0003-0.0002-0.00010

compression reinforcement strain

Load

(kN

)

strain gage 3strain gage 4


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FIGURE 11 Load versus compressive strain in the concrete

FIGURE 12 In-situ bridge load test

0

25

50

75

100

125

150

175

200

225

-0.0025-0.002-0.0015-0.001-0.00050concrete strain

Load

(kN

)

strain gage 9strain gage 10strain gage 11


Panel 1 2 3 4 5 6 7 8 9

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FIGURE 13 DCVT transducer locations

FIGURE 14 Lateral location of truck Passes 1 th

N

N

Pass 4

Pass 3

Pass 2

Pass 1

Mid-span

ro
Directionof
Traffic

Panel 1 2 3 4 5 6 7 8 9

ugh 4

Direction of

Traffic


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FIGURE 15 Deflected shape of the bridge for Pass 1


0

0.5

1

1.5

2

2.5

3

0 1 2 3 4 5 6 7

Lateral location on the bridge (m)

Def

lect

ion

(mm

)

Stop 1Stop 2Stop 3Stop 4Stop 5

0

0.5

1

1.5

2

2.5

3

0 1 2 3 4 5 6 7Lateral location on the bridge (m)

Def

lect

ion

(mm

)



20


0

0.5

1

1.5

2

2.5

3

0 1 2 3 4 5 6 7


Def

lect

ion

(mm

)

Stop 1Stop 2Stop 3Stop 4Stop 520 mph


0

0.5

1

1.5

2

2.5

3

0 1 2 3 4 5 6 7Lateral location on the bridge (m)

Def

lect

ion

(mm

)



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0

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0.04

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0.07

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0.09

0.1

1 2 3 4 5 6 7 8 9

Panel Number

Perc

enta

ge o

f Tot

al L

oad

Pass 1

Pass 2

Pass 3

Pass 4

FIGURE 19 Percentage of load carried per panel as a percentage of total

load on the bridge – Passes 1 through 4

0

0.01

0.02

0.03

0.04

0.05

0.06

0.07

0.08

1 2 3 4 5 6 7 8 9

Panel Number

Perc

enta

ge o

f Tot

al L

oad

Pass 4

32kph (20mph)

FIGURE 20 Percentage of load carried per panel as a percentage of total

load on the bridge – Pass 4 and 32khp (20mph) pass


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FIGURE 21 Superposition of two truck loads on the bridge. Table 1 Results of the tensile characterization of the FRP bars.

Failure Stress, MPa (ksi)

Modulus of Elasticity, GPa (Msi)

CFRP 9.5 mm (3/8 in) 2127 (308) 109.2 (15842)

2183 (317) 120.0 (17409) Mean 2155 (312.5) 114.6 (16625.5)

Standard Deviation 39.6 (6.4) 7.6 (1108 .4) Variability 1.9% 6.7%

GFRP

9.5 mm (3/8 in) 889 (129) 36.9 (5358) 857 (124) 37.4 (5442)

Mean 873 (126.5) 37.2 (5400) Standard Deviation 22.6 (3.5) 3.3 (59.4)

Variability 2.7% 1.0%

GFRP 12.7 mm (1/2 in) 789 (114) 35.6 (5167)

789 (114) 36.6 (5311) Mean 789 (114) 36.1 (5239)

Standard Deviation 0 0.7 (101.8) Variability 0.0% 1.9%

0

0.5

1

1.5

2

2.5

3

3.5

0 1 2 3 4 5 6 7


Def

lect

ion

(mm

)



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Table 2 Comparison of laboratory results and manufacturer’s recommended values.

Failure Stress Modulus of Elasticity Manufacturer

Data Laboratory

Data Difference Manufacturer

Data Laboratory

Data Difference

CFRP 9.5 mm (3/8 in)

1862 MPa (270ksi)

2155 MPa (312.5 ksi)

+15.7% 104.8 GPa (15.2 Msi)

114.6 Gpa (16.6 Msi)

+9.2%

GFRP 9.5 mm (3/8 in)

779.1 MPa (113 ksi)

873 MPa (126.5 ksi)

+12.0% 41.4 GPa (6 Msi)

37.2 GPa (5.4 Msi)

-10.0%

GFRP 12.7 mm (1/2 in)

723 MPa (105 ksi)

798 MPa (114 ksi)

+10.4% 41.4 GPa (6 Msi)

36.1 GPa (5.24 Msi)

-12.5%

Table 3 Truck Axle Spacing. Center-to-center spacing

(m) Out-to-out spacing

(m) WIDTH Front axle 2.02 2.29 Middle axle 1.87 2.41 Rear axle 1.87 2.41 LENGTH Front axle to Middle axle 4.60 Middle axle to Rear axle 1.35 Table 4 Longitudinal Truck Locations. Stop Truck Position 1 Middle and rear axles of the truck centered

approximately 0.30 m (1 ft) onto the bridge from the north end

2 Middle and rear axles of the truck centered approximately 1.83m (6 ft) onto the bridge from the north end

3 Middle and rear axles of the truck centered approximately 3.35 m (11 ft) onto the bridge from the north end (i.e., at midspan)

4 Middle and rear axles of the truck centered approximately 4.88 m (16 ft) onto the bridge from the north end

5 Middle and rear axles of the truck centered approximately 6.40 m (21 ft) onto the bridge from the north end

deflection assessment of an frp-reinforced concrete...

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