estimated versus measured capacity of cfa piles for

9h Australian Small Bridges Conference 2019

Estimated versus measured capacity of CFA piles for Seaford Road

bridges, Melbourne

Cillian Mc Colgan, Associate Geotechnical Engineer, WSP

Suthagaran Visvalingam, Senior Geotechnical Engineer, WSP

Nicholas Withers, Geotechnical Engineer, WSP

ABSTRACT

This paper discusses the design of piles for a rail bridge in south east Melbourne which adopted a

relatively conventional approach to estimating static axial pile capacity and compares it to construction

stage validation.

On site validation included assessment of pile installation data provided by the piling rigs on board

computer. Some agreement between the interpreted ground profile and the data obtained from the

piling rig outputs was observed.

The results of fifteen dynamic pile tests and two sacrificial (non-working) pile tests are also presented.

These tests demonstrated greater capacity than may be estimated through these conventional

approaches. This presents opportunity to further refine pile design on subsequent stages of the

project leading to a more robust and cost effective design.

1 INTRODUCTION

The removal of level crossings improves safety by separating trains from traffic, with other benefits

such as reduced congestion, an upgraded road network, connected communities and support of

urban regeneration. The current Victorian State Government has initiated a level crossing removal

program to remove at least 75 of these level crossings across Melbourne by 2025. The Southern

Program Alliance (SPA) comprising WSP, Lendlease, Acciona/Coleman Rail, MTM and the Level

Crossing Removal Project (LXRP) is delivering the works associated with the Frankston Line which

includes the removal of four level crossings at Seaford Road in Seaford and at Mascot Avenue,

Bonbeach and Station Street and Eel Race Road in Carrum. The site location is shown in Figure 1.

Figure 1- Seaford Road site location (LXRP, https://levelcrossings.vic.gov.au/projects )

https://levelcrossings.vic.gov.au/projects

https://levelcrossings.vic.gov.au/projects


At Seaford in south east Melbourne, two parallel U trough girder bridges have been constructed with

combined spans of more than 110m. These bridges convey the Frankston Line over a shared user

path and Seaford Road. These bridges are founded on groups of Continuous Flight Auger (CFA)

piles.

A photograph of the completed bridge is reproduced at Figure 2 below.

Figure 2- Seaford Road bridge (LXRP, Image Gallery of Seaford Road, Seaford)

2 GEOTECHNICAL MODEL

The ground conditions at Seaford consist of surficial fill, between 3 m to 5 m of quaternary aged

sands and swamp deposits overlying Tertiary aged Brighton Group (now officially known as the

Sandringham Sandstone) and Gellibrand Marl Units These conditions are typical of South East

Melbourne.

The salient features of these materials, as they relate to the Seaford bridges are summarized in Table

1. The geological long section and a detail of the Southern Abutment of the Seaford Road Bridge are

presented in Figure 2 and Figure 3.

Older Volcanics are also encountered at Seaford but these underlie the Gellibrand Marl and were not

considered in the bridge foundation design due to their depth.

Table 1: Geotechnical units relevant to Seaford Road Bridge

Geological Unit and ID Material Description Consistency Depth to top (m bgl)

Unit 1 Fill Shallow fill associated with

rail formation

N/A 0.01

Unit 2 2A Quaternary

Sands

Dune Sands Medium

Dense

0.0 – 1.0

2B Swamp

deposits

Sands with occasional layers

of compressible back swamp

materials

Loose 2.9 – 6.51

Unit 3 3A Brighton Group Sandy Clay Stiff to Hard 4.1 – 8.0


3B (Sandringham

Sandstone)

Clayey Sand, occasionally

cemented

Medium

Dense to

Very Dense

10.0 – 16.5

Unit 4 Gellibrand Marl Stiff to Very Stiff Stiff to Very

Stiff

19.0 – 22.6

Notes:

1. Unit not present across all pier/abutment locations

Figure 3- Geological long section for the Seaford Bridge, Seaford

3 AXIAL CAPACITY ESTIMATION AND PILE DESIGN

Table 2 presents the adopted design parameter ranges for the different units including the ranges of

Standard Penetration Test (SPT) “N” values recorded in the geotechnical investigations.

Table 2: Geotechnical design parameters and SPT results

Geological Unit ID

Adopted SPT “N” Value

Adopted Design Parameters

Undrained

Shear

Strength, Su

(kPa)

Effective

Cohesion, c’

(kPa)

Effective

Friction

Angle, ’

(Degrees)

Ultimate

Unit Skin

Friction, fs

(kPa)

Ultimate Unit

End Bearing

Friction, fb

(kPa)

Unit 2A 16 N/A 0 29° – 32° 15 - 24 N/A

Unit 2B 0 - 23 N/A 0 28° – 33° 4 - 32 N/A

Unit 3A 12 - 33 40 - 200 - - 16 - 65 1100 - 3600

Unit 3B 12 - 33 N/A 0 30° – 36° 16 - 65 1100 - 3600

Unit 4 14 - 25 125 - 175 0 28° – 33° 39 - 43 700 – 1200


Shaft friction and end bearing was limited to the values recommended by Decourt, L., (1995). The

average skin friction was assessed based on O’Neill & Reese (1999) for cohesive soils and Craig

(2004) for granular soils. The ultimate bearing capacity was assessed based on Fleming et. al (2009)

for cohesive soils and Berezantzev (1961) for granular soils.

Based on the above interpretation of the ground conditions and the provided loading, pile lengths

were assessed and are presented in Table 3 below. Note that the design geotechnical strength is

provided for each pile which was based on a geotechnical strength reduction factor (g) of 0.72. This

was adopted based on undertaking a dynamic load test at each pier and abutment location (i.e. 8

piles out of a total of 80 piles for the bridge foundation). The strength presented in Table 3 is also the

load that needed to be achieved in the dynamic testing.

Table 3: Pile design summary

Bridge ID Foundation Location

Pile Diameter

(mm)

Required Ultimate Capacity at Pile Head1

(kN)

Pile Length (m)

Founding Material

Pedestrian

Access Bridge

Abutment A 1050 3740 20.0 Stiff Clay

Abutment B 1050 3740 19.5 Very Stiff Clay

Seaford Road

Bridge

Abutment C 1050 3740 19.0 Very Stiff Clay

Pier 1A/1B 1050 3650 18.0 Very Stiff to Hard Clay

Pier 2A/2B 1050 3650 19.5 Stiff to Very Stiff Clay

Abutment D 1050 3740 19.0 Very Stiff Clay

Notes:

1. This does not include pile self weight and is also the pile test load

2. The capacity of the pile is measured after the pile is cast so the pile self weight during the

test, thus this can be added to the result to get the piles real capacity

4 CONSTRUCTION VALIDATION

The primary method of validation of the design was through testing piles at each bridge support

location in accordance with the requirements of VicRoads Specification 607. Periodic inspections of

the drilling operations were also carried out along with a review of the pile installation records

provided by the Continuous Flight Auger (CFA) piling rig.

4.1 Pile installation records

The piles were installed using a LB28 piling rig with a Jean Lutz on board computer. The onboard

computer measures a number of different operation parameters including drill pressure, intrusion rate

and pile radius. The latter is calculated by comparing the volume of concrete delivered with the rate of

extraction of the auger.

An example of the computer outputs compared to a borehole log is reproduced in Figure 4. Although

somewhat anecdotal, there does not appear to be a very convincing correlation between drill pressure

and material type. This was found to be consistent across all the pile installation records.

It is the opinion of the author that this data does not provide a reliable means of validating ground

conditions for floating piles. The assessment undertaken was focused on verifying that conditions

were not significantly worse than assumed in the design, i.e. no significant drop off in drill pressures

were observed and a general trend of drill pressure increasing with depth was present and the as built

pile diameter achieved was in accordance with the design.


Figure 4- Pile installation records compared to adopted geotechnical ground model

4.2 Inspection of Pile Cuttings

Inspecting the pile cuttings was carried out at a nominal frequency of about 2 piles per group. A plan

showing the pile caps for Piers 2A and 2B and Abutment D is reproduced at Figure 5 below.

Figure 5- Plan illustrating pile layouts

Pier 2A/B

Abutment D


There are technical difficulties associated with trying to assess the materials a CFA pile is founded in

by logging auger cuttings. These include:

• mixing of soils during drilling making it difficult to assess what level the soils on the CFA auger

come from

• disturbance of soils during drilling affecting their consistency, especially for the sandy silts which

were the soils this bridge was founded in.

There is also a logistical disadvantage to not knowing the founding material you are in until

completion of pouring the pile. Its essentially too late to do anything other than re-excavation of the

already poured pile.

Where floating CFA piles are proposed geotechnical investigations should be carried out at an

appropriate density. Relying on validating the materials during construction is not practical.

Complementing the investigations with pile testing is a more robust way of validating the design and

pays dividends through increased φg. and leading to a more robust and cost effective design.

4.3 Dynamic Pile Testing – Proof Tests

A purpose-built test frame equipped with a 12T drop hammer was used to test the nominated piles.

Drop heights of between 0.8 and 1.5 m were used which equates to a test energies of between 94

and 176 kJ.

Table 4 below shows the results of the pile testing versus the required test load. Capacities far in

excess of that required by the design were achieved. All piles demonstrated the required test loads on

the first attempt. Piles were left to set up for a minimum of 6 days prior to test.

The capacities demonstrated below were done so at relatively modest test energies. The testing was

not specified to demonstrate the full capacity of the piles, rather the required capacity.

Table 4: Dynamic pile load test results of Seaford Bridge foundations

Bridge Foundation Location

Estimated Pile Capacity1

(kN)

Measured Pile Capacity2 (kN)

Estimated Shaft Force (kN)

Measured

Shaft

Force

(kN)

Estimated

End

Bearing

Force

(kN)

Measured

End

Bearing

Force

(kN)

Abutment A - P09 3746 9575 3221 6112 925 3463

Abutment B - P04 3751 6487 2934 5188 1195 1299

Abutment C - P03 3741 9303 2764 6273 1360 3030

Pier 1A - P01 3651 8869 2653 7138 1359 1731

Pier 1B - P01 3651 7746 2653 6231 1359 1515

Pier 2A - P02 3641 7513 3206 5781 820 1731

Pier 2B - PP08 3641 7336 3206 5605 820 1731

Abutment D –

P04

3745 7323 1927 5592 2098 1731

Notes:

1. The proportion of the piles capacity used up by self weight has been omitted for ease of

comparison to the pile test result


2. Measured capacity above pile cut off level has been subtracted from the overall test result

3. The capacity of the pile is measured after the pile is cast so the pile self weight is active

during the test, thus this can be added to the result to get the piles real capacity

4. Pile diameter for all the bridge foundation is 1050 mm

5. Pile length for the bridge foundation location is as per the Table 3

4.4 Dynamic Pile Testing – CFA rigid inclusions

The bridge approach embankments were founded above potentially compressible soils and as such

some ground improvement was required to manage settlement risk. This took the form of unreinforced

“rigid inclusions” constructed as unreinforced CFA piles. Like the CFA piles these demonstrated much

higher strengths than would be calculated through conventional means. These were tested with the

same apparatus but at lower energies than the CFA piles. Test results are presented in Table 5.

Table 5: Dynamic pile load test results of CFA rigid inclusion

Rigid Inclusions Location

Measured Pile Capacity1

(kN)

Diameter

(mm)

Length

(m)

Measured

Shaft Force

(kN)

Estimated End

Bearing Force

(kN)

AD87 2427 500 8.6 1691 736

AB32 1592 500 8.6 840 742

AA06 1989 500 10.6 1351 638

PN59 1200 500 8.4 660 540

PN70 4879 600 17.0 3041 1838

PN61 4520 600 17.0 3177 1343

PN55 5641 600 17.0 3945 1696

Notes:

1. Measured capacity includes working platform material

4.5 Dynamic Pile Testing – Destructive Tests

Before the completion of the pile testing it was recognized that significantly higher pile capacities than

assumed in the design were being achieved. Also, the tests on the rigid inclusions were succeeding in

demonstrating relatively high end bearing at shallower depths. Two destructive tests were undertaken

at Seaford to try to investigate if even higher capacities could be demonstrated. This investment

(approximately $50,000) was justified against the potential savings that could be realized on a bridge

to be designed in similar materials later that year.

Table 6: Dynamic pile load test results of destructive tests

Location Diameter

(mm)

Length

(m) Measured Capacity1

(kN)

Measured

Shaft Force

(kN)

Measured End

Bearing Force

(kN)

Test Pile

Abutment A2

1050 9.0 3472 1740 1732

Test Pile

Abutment C3

1050 14.0 9040 3412 5628

Notes:

1. Measured capacity includes working platform material

2. Pile for destructive test at Abutment A is founded in Unit 3A Brighton Group Sand

3. Pile for destructive test at Abutment C is founded in Unit 3B Gellibrand Marl


5 ANALYSIS OF PILE TEST RESULTS

5.1 Shaft Friction

The results of the 17 dynamic tests have been compared to SPT testing results to see if the results

correlate. A summary of all of the dynamic pile load test data, for each geological units along with the

SPT test results are presented at Figure 4. The SPT tests were corrected for energy and overburden

stress (for granular soils) using the methodology presented by Skempton (1996).

Table 7 compares the average SPTN1,60 values for each of the units with the measured shaft friction.

Table 7: Summary of shaft friction

Unit N1,60 Shaft Friction (kPa) Correlation

Range Average Range Average

Unit 2 0 – 43 15 50 – 80 65 Fs = 4.3*N1,60

Unit 3A 10 – 71 23 50 - 120 80 Fs = 3.5*N1,60

Unit 3B 0 - 70 33 80 - 150 100 Fs = 3.0*N1,60

Unit 4 10 – 65 21 50 - 150 100 Fs = 4.7*N1,60

Broadly speaking there does appear to be good correlation between N1,60 and the measured shaft

friction in the pile tests for the Brighton Group Materials (Unit 3A and 3B) and the Quaternary Sands

(Unit 2). A general trend of increasing shaft friction and N1,60 with depth is clear.

The Gellibrand Marl (Unit 4) doesn’t follow this trend. It is difficult to assess exactly why the data for

the Gellibrand Marl shows such high shaft friction. The Gellibrand Marl was generally logged as

cohesive and the undrained strengths recorded were relatively low when compared to the measured

shaft friction. It could be to do with the use of the SPT as the primary form of characterizing these

soils. Silts, especially when they are sandy like we found at Seaford, tend to dilate. It is easy to

imagine this leading to the development of local excess pore pressures when they are subjected to

the first blow from an SPT. This could be responsible for the relatively low values of N1,60.

The pile test however occurs some 6 days after the pile has been set. This provides time for excess

pore pressures that may have been generated during pile installation to dissipate and for the silts to

move towards there equilibrium condition. Allowing piles to set up before testing is a common

practice. We do not do this, however, when conducting in-situ testing on soils that may be subject to

dilation.

The outcome of the above is two fold, the correlations may not be appropriate for the Gellibrand Marl

given doubts around the applicability of SPT testing. Other, slower (and more expensive) means of

testing in-situ strength may be better capable of characterizing the Gellibrand Marl but these are

expensive and come with their own challenges.


Figure 6- Pile test results compared to SPT N1,60

0 2,000 4,000 6,000 8,000

-22

-20

-18

-16

-14

-12

-10

-8

-6

-4

-2

0

0 40 80 120 160 200

Unit End Bearing (kPa)

RL

(m A

HD

)

Unit Skin Friction (kPa)

Unit 2 Unit 3A Unit 3B Unit 4

-22

-20

-18

-16

-14

-12

-10

-8

-6

-4

-2

0

0 10 20 30 40 50 60

RL

(m A

HD

)

SPT N1,60

Unit 2A Unit 2C Unit 3A

Unit 3B Unit 4


5.2 End Bearing

Table 8 compares the average SPTN1,60 values for each unit with the measured end bearing values.

Table 8: Summary of end bearing

Pile Location ID Geological Unit

SPT N1,60 at pile toe1

Measured End Bearing, Fb (kN)

Correlation Between Fb and SPT N1,60

Test Pile at Abutment C Unit 3A 26 2000 Fb = 77 x N1,60

AD87 Unit 3A 18 3750 Fb = 208 x N1,60

AB32 Unit 3A 18 3750 Fb = 208 x N1,60

AA06 Unit 3A 68 3680 Fb = 54 x N1,60

PN59 Unit 3A 27 6000 Fb = 222 x N1,60

Averages for Unit 3A 33 4295 Fb = 131 x N1,60

Test pile Abutment A Unit 3B 29 6500 Fb = 224 x N1,60

PN70 Unit 3B 30 6500 Fb = 217 x N1,60

PN61 Unit 3B 30 4750 Fb = 158 x N1,60

PN55 Unit 3B 30 6000 Fb = 200 x N1,60

Averages for Unit 3B 30 5938 Fb = 200 x N1,60

Abutment A - P09 Unit 4 15 4000 Fb = 267 x N1,60

Abutment B - P04 Unit 4 18 1500 Fb = 83 x N1,60

Abutment C - P03 Unit 4 22 3500 Fb = 159 x N1,60

Pier 1A – P01 Unit 4 36 2000 Fb = 56 x N1,60

Pier 1B – P01 Unit 4 36 1750 Fb = 49 x N1,60

Pier 2A – P02 Unit 4 13 2000 Fb = 154 x N1,60

Pier 2B – PP08 Unit 4 13 2000 Fb = 154 x N1,60

Abutment D Unit 4 16 2000 Fb = 125 x N1,60

Averages for Unit 4 21 2344 Fb = 111 x N1,60

Notes: 1. Nominally includes SPT tests 2 diameters above and 3 diameters below toe

There is a considerable amount of scatter in the data. This is likely due to the variation in energy

during testing and the ranges of pile sizes considered. The smaller piles founded in Units 3A and 3B

from CMC’s show higher end bearing than the larger piles in Gellibrand Marl. This is more likely

associated with the difficulty of mobilizing end bearing in large and long end bearing piles than a

difference in strength. It is logical to expect that more end bearing could be available in the Gellibrand

Marl based on the relatively high amounts of shaft friction mobilized.

The destructive test at Abutment A does show that these capacities are possible for larger piles.

5.3 Test Energy

Figures 6 to 9 present the range of measured skin friction versus test energy. It appears from the data

that the test energies were generally sufficient to mobilise all the available shaft friction for the piles

tested in the upper materials. The Gellibrand Marl may be capable of demonstrating more shaft

friction at higher test energies as suggested by Figure 9.


Figure 7- Test energy versus measured skin friction – Unit 2

Figure 8- Test energy versus measured skin friction – Unit 3A

20

30

40

50

60

70

80

90

100

110

120

130

140

150

160

0 50 100 150 200 250 300 350 400

Mea

sure

d S

kin

Fri

ctio

n (

kPa)

Applied Energy (kJ)

Unit 2 - Quaternary Sand & Swamp Deposits

AA-09 - U2 AB-04 - U2 AC-03 - U2 AD-04 - U2 CMC-AA-06 - U2 CMC-AB-32 - U2CMC-PN-59 - U2 CMC-AD-87 - U2 CMC-PN-55 - U2 CMC-PN-61 - U2 CMC-PN-70 - U2 P1-A1 - U2

20

30

40

50

60

70

80

90

100

110

120

130

140

150

160

0 50 100 150 200 250 300 350 400

Mea

sure

d S

kin

Fri

ctio

n (

kPa)

Applied Energy (kJ)

Unit 3A - Brighton Group - Sandy Clay

AA-09 - U3A AB-04 - U3A AC-03 - U3A AD-04 - U3A CMC-AA-06 - U3ACMC-AD-87 - U3A CMC-PN-55 - U3A CMC-PN-61 - U3A CMC-PN-70 - U3A P1-A1 - U3A

2 results

2 results

3 results


Figure 9- Test energy versus measured skin friction – Unit 3B

Figure 10- Test energy versus measured skin friction – Unit 4

20

30

40

50

60

70

80

90

100

110

120

130

140

150

160

0 50 100 150 200 250 300 350 400

Mea

sure

d S

kin

Fri

ctio

n (

kPa)

Applied Energy (kJ)

Unit 3B - Brighton Group - Clayey Sand

AB-04 - U3B AA-09 - U3B AC-03 - U3B AD-04 - U3B

CMC-PN-55 - U3B CMC-PN-61 - U3B CMC-PN-70 - U3B P1-A1 - U3B

P1-B1 - U3B P2-A2 - U3B P2-B8 - U3B TP-AA - U3B

20

30

40

50

60

70

80

90

100

110

120

130

140

150

160

0 50 100 150 200 250 300 350 400

Mea

sure

d S

kin

Fri

ctio

n (

kPa)

Applied Energy (kJ)

Unit 4 - Gellibrand Marl

CMC-PN-55 - U2 CMC-PN-61 - U2 CMC-PN-70 - U2 P1-A1 - U2 P1-B1 - U2

P2-A2 - U2 P2-B8 - U2 AA-09 - U4 AC-03 - U4 AD-04 - U4

2 results

3 results

2 results


Figures 10 to 14 show the relationship between the applied test energy and the mobilized end bearing

resistance for the different units encountered at Seaford. There is a clear trend of an increase in the

measured end bearing with an increase in the test energy. It can also be seen that the smaller

diameter piles (CFA Rigid Inclusions) that were tested we capable of demonstrating more end bearing

at lower energy, as would be expected.

Figure 11- Test energy versus measured end bearing for all units

Figure 12- Test energy versus measured end bearing – Unit 4

0

500

1000

1500

2000

2500

3000

3500

4000

4500

5000

5500

6000

6500

7000

0 50 100 150 200 250 300 350 400Mea

sure

d E

nd

Bea

rin

g(kP

a)

Applied Energy (kJ)

All Units

AA-09 - U4 AB-04 - U4 AC-03 - U4 AD-04 - U4 P1-A1- U4

P1-B1 - U4 P2-A2 - U4 P2-B8 - U4 TP-AA - U3B TP-AC - U3A

CMC-AA-06 - U3A CMC-AD-87 - U3A CMC-PN-55 - U3B CMC-PN-61 - U3B CMC-PN-70 - U3B

0

500

1000

1500

2000

2500

3000

3500

4000

4500

5000

5500

6000

6500

7000

0 50 100 150 200 250 300 350 400

Mer

asu

red

En

d B

eari

ng(

kPa)

Applied Energy (kJ)

Unit 4 - Gellibrand Marl

AA-09 - U4 AB-04 - U4 AC-03 - U4 AD-04 - U4 P1-A1- U4 P1-B1 - U4 P2-A2 - U4 P2-B8 - U4

Rigid Inclusions


6 CONCLUSIONS

In summary the pile test data has shown:

- Conventional design methods can be overly conservative for pile design in the Brighton

Group and Gellibrand Marl Units

- The Brighton Group and Gellibrand Marl Units both displayed skin friction far greater than

conventional design would suggest.

- The Brighton Group units showed a similar trend for end bearing though the data to support

this is not as comprehensive as that for shaft friction and it is limited to smaller diameter piles

- There is likely more end bearing capacity available in the Gellibrand Marl than was

encountered at Seaford bridge

- Potential to prove greater capacities with destructive testing, which may not be limited to

particular geological units.

The results clearly demonstrate that conventional pile design methods are conservative for the

Brighton Group and Gellibrand Marl units. The difference between the test results and the

conventional design approach is striking and should encourage geotechnical engineers to pursue pile

testing more aggressively. It is the authors opinion that in the long term this will result in a net saving

for industry, especially for major infrastructure projects.

The Southern Program Alliance is currently pursuing the aforementioned opportunity with the design

of a larger bridge founded in similar materials and has been able to use the data obtained at Seaford

to achieve a more economical foundation design.

7 ACKNOWLEDGEMENTS

The author would like to thank the Level Crossing Removal Project and the Southern Program

Alliance for permission to present this work.

8 REFERENCES

Berezantzav, V.G. et. al. (1961), Load bearing capacity and deformation of piled foundations, in

Proceedings 5th International Conference SMFE, Paris, Vol. 2, pp. 11-15.

Craig, R.F. (2004) Craig’s Soil Mechanics, 7th edition, Spon Press, Taylor & Francis Group, New

York, USA, pp. 313-315.

Clarke and Leonard (2004), Regional variations in neo-tectonic fault behaviour in Australia, as they

pertain to the seismic hazard in capital cities, Australian Earthquake Engineering Society 2014

Conference, Nov 21-23, Lorne, Vic.

Fleming W.G.K. (1992), A New Method for Single Pile Settlement Prediction and Analysis,

Geotechnique 42, No. 3, pp. 411-425.

Fleming, K. et. al. (2009), Piling Engineering, 3rd edition, Taylor & Francis Group, New York, USA, pp.

108.

O’Neill, M.W. and Reese, L.C. (1999), Drilled Shafts: Construction Procedures and Design Methods,

Publication No. FHWA-IF-99-025, Federal Highway Administration, Washington, D.C., pp 758.

Poulos, H.G and Davis, E.H. (1980) Pile Foundation Analysis and Design, Rainbow-Bridge Book Co,

Canada, pp. 31.

Decourt L. (1995), Prediction of load - settlement relationship for foundations on the basis of the SPT,

Ciclo de Conferencias International, Leonardo Zeevaert, UNAM, Mexico, pp.85-104.

Skemption, A. W., Standard penetration test procedures and the effects in sands of overburden

pressure, relative density, particle size, ageing and overconsolidation. Geotechnique 36, No 3, 425 -

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estimated versus measured capacity of cfa piles for

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