geotechnical design recommendations for pedestrian bridge ...€¦ · the pedestrian bridge will be...

June 17, 2013 File: 17-123-786 Associated Engineering (B.C.) Ltd. Suite 300 - 4940 Canada Way Burnaby, B.C. V5G 4M5 Attention: Mr. Craig Schaper, P.Eng.

GEOTECHNICAL DESIGN RECOMMENDATIONS FOR PEDESTRIAN BRIDGE FROM 4TH STREET PARKADE TO WATERFRONT PARK

NEW WESTMINSTER, B.C. Dear Craig: As requested, Thurber Engineering Ltd. (TEL) has completed a geotechnical assessment for a proposed pedestrian bridge linking the Front Street Parkade to Waterfront Park at 4th Street. This report presents the results of our own assessment and our geotechnical recommendations for foundation design. This report was prepared for the exclusive use of Associated Engineering (B.C.) Ltd. (AE), and the City of New Westminster (City). Any use that a third party makes of this letter, or any reliance on decisions based on it are the responsibility of such third parties. TEL accepts no responsibility for damage incurred by third parties as a result of decisions made or actions taken based on this report. Use of this report is subject to the attached Statement of Limitations and Conditions that are an integral part of this report. 1. INTRODUCTION We understand that the City plans to construct a pedestrian bridge over the existing road and railroad tracks in order to provide access to the recently completed waterfront park at the south end of 4th Street. The pedestrian bridge will be a free-standing cantilever structure supported on seven, 914 x 15.9 mm steel pipe piles with a minimum yield stress of 310 MPa. According to AE, the bridge will not be structurally connected to the parkade structure. The structure will be designed in accordance with the CAN/CSA-S6-06 (bridge code), which considers a design seismic event with 10% chance of exceedence in 50 years (1:475). Our scope of work on this project consists of providing geotechnical input to aid in the design of the bridge foundation based on available information. Assessment of geotechnical work carried out during the construction of the park, and environmental considerations is not within the scope of work.

900, 1281 West Georgia Street, Vancouver, BC V6E 3J7 T: 604 684 4384 F: 604 684 5124thurber.ca

Client: Associated Engineering (B.C.) Ltd. Date: June 17, 2013 File No.: 17-123-786 E-File: c_mcb_RPT_4th Street Pedestrian Bridge 2013-06-17 of 10

2. REFERENCE DOCUMENTS Our understanding of the proposed development is based on the documents and drawings listed below. For discussion purposes we will refer to the document numbers shown here.

1. Trow Associates Inc., April 9, 2010. Geotechnical Factual Report and Seismic Analysis, Preliminary Recommendations at 30% Design Stage for Project Master Plan.

2. Trow Associates Inc., June 25, 2010. Geotechnical Recommendations Update New Westminster Waterfront Park.

3. GeoPacific Consultants Ltd., May 17, 2011. New Westminster Pier Park – Delta-Lok Retaining Wall, G-W1 through G-W9.

4. Worley Parsons, April 13, 2013. Westminster Pier Park Structural Building Concession / Washroom Building Plan and Sections, 09216-00-ST-DRD-1126 Rev.2.

5. Associated Engineering, February 19, 2013. 4th Street Pedestrian Overpass – Structural General Arrangement – 2297-103 Rev. B and 2297-107 Rev.A

3. BACKGROUND The design of the bridge foundation is particularly challenging from a geotechnical standpoint due to the development history of the site. The waterfront park area was reclaimed from the Fraser River in the late 19th century using random fill (mineral, construction debris, occasional wood waste). The fill materials were generally end dumped with no compaction. As such, the fill soils are relatively loose and would be considered highly susceptible to liquefaction, with a consequence of liquefaction being large lateral displacements (i.e. flow slide) into the Fraser River. As part of the waterfront park construction, timber densification piles were driven through the fill to practical refusal in the native, dense till-like soils, on-land and off-shore, to improve the resistance to liquefaction of the fill soil with the intent of reducing the potential lateral displacements associated with the design level earthquake. The off-shore timber densification was designed to accommodate the 1:475 year event level and piles were installed at a 1.2 m square spacing, whereas the on-shore piles were designed to accommodate the 1:2475 year event level and installed at a 1.25 by 1.05 m rectangular spacing. However, the extent of the on-shore densification was limited to within about 10 m of an existing steel bulkhead, and at building locations. The sheet pile wall is restrained by a buried concrete deadman located outside of the area of the proposed bridge foundation. Prior to the development of the park the typical site grade was at El. +4 m geodetic. During park construction site grades along the landward property line were raised as high as El. +8 m while site grades near the steel bulkhead remained at about El. +4 m. The grade increase near


the landward property line was accomplished using a 4 m high Delta-Lok retaining wall. Document 3 shows that the existing Delta-Lok wall is reinforced with geogrid (Stratagrid SG200) with lengths equal to 0.7 of the embankment height (or about 3 m) and a vertical spacing of 0.3 m or two filled Delta-Lok GTX-M soils bags (whichever is less). No ground improvement was carried out along the wall. As such, during a seismic event the existing wall would be expected to displace laterally towards the tracks. It should be noted that this movement does not appear to have been addressed in the provided information concerning the development of the park. A concession building was constructed a short distance downriver (about 10 m) of where a projection of 4th Street would enter the park. To mitigate building settlement light weight fill consisting of expanded polystyrene (EPS) was used adjacent to the building. Document 4 shows the layout of the EPS in plan and section. In plan, the EPS extends about 5 m towards the landward property line and 10 m upriver of the building. However, there is a discrepancy as the edge of the EPS is about 6.5 m behind the toe of the MSE retaining wall in the cross section and approximately 5 m in plan. The proposed bridge foundation is to be installed behind the existing Delta-Lok retaining wall, oriented along the centreline projection of 4th Street. We anticipate that that the four landward piles will be installed within the zone of geogrid reinforced soil, and that the three riverward piles may be installed within the EPS zone. 4. SUBSURFACE CONDITIONS 4.1 General No site investigation has been carried out by TEL in preparation of this report. Our understanding of the subsurface conditions is based on the past site investigation work by Trow Associates Inc (Document 1). It should be noted that electronic records of cone penetration tests (CPT) performed during the 2010 site investigation were provided. 4.2 Inferred Subsurface Conditions An inferred generalised stratigraphy of the subsoils at the bridge location, based on the information provided, is summarised in Table 1 below. A selection of relevant test hole logs is attached. These logs form the basis of our geotechnical assessment and design.


Table 1 – Inferred Subsoil Profile at Bridge Foundation Location

Soil Unit Thickness

(m) Elevation

(m) Description

4 +8 to +4 Granular Fill for MSE Wall

A 3 +4 to +1 Granular Fill

compact to loose, SAND with varying gravel content, trace to no silt; above water table

B 5 +1 to -4

Upper Liquefiable Granular Fill loose to compact, SAND some gravel to sandy

SILT / silty SAND, occasional wood/construction debris

C 4 -4 to -8 Lower Liquefiable Granular Fill

loose to compact, sandy SILT / silty SAND, occasional wood/construction debris

D 6 -8 to -16 Till-Like

dense to very dense, grey silty gravelly SAND

E to bottom below -16

Hard Silt very stiff to hard, grey SILT some clay, occasional

sand seams / pockets, occasional gravel; at or near plastic limit, 18 < w(%) < 30

The native, dense till-like and underlying hard silt would be considered to have a low to negligible susceptibility to liquefaction. Subsoil profile sections C-C’ and D-D’ from Document 1, and located approximately 100 m downstream and 25 m upstream from the bridge location, respectively, are attached as Figures 1 and 2. 4.3 Groundwater Document 1 indicates that the groundwater levels measured to the north of the bulkhead varied between El. 0 to +2.0 m. The ground water level at the bridge location would be expected to vary seasonally with the river level, and daily with the tides. An average ground water level of El. +1.0 m (depth of 3 m) was assumed for our design. 5. SEISMIC ASSESSMENT 5.1 Site Specific Response Analysis (SSRA) A SSRA was performed using the program EERA (equivalent-linear earthquake response analysis). EERA is a modern spreadsheet implementation of the well-known concepts of equivalent linear earthquake analysis used in programs SHAKE. The shear wave velocity values used in the analysis were taken from Document 1. We understand that the shear wave


velocity measurements were not obtained from the actual waterfront park site, but rather the (former) Riverview Casino parking lot located approximately 500 m downstream. The stratigraphy used in the analysis was based on nearby test holes. The results of the SSRA are shown on the attached Figure 3. For comparison purposes, the response spectra interpreted from the bridge code are also included. The bridge code response spectra were estimated assuming Soil Class III, Importance Factor 1.0, and a peak horizontal acceleration (PHA) of 0.267 g in accordance with the NBC 2010 seismic hazard value interpolation website. It should be noted that the calculated SSRA response spectrum exhibits amplification above the bridge code response spectra between about 0.4 and 0.8 seconds. Based on the results of the SSRA, the design response spectrum should be the greater of SSRA or bridge code for any given period. 5.2 Liquefaction Assessment The liquefaction assessment was carried out using the procedures presented in Youd et al. (2001) and Idriss & Boulanger (2008) for CPT and SPT data, respectively. The cyclic stress ratio (CSR) used in the assessment was obtained from the SSRA. Post liquefaction settlement was assessed using the procedures presented in Zhang et al. (2002) and Idriss and Boulanger (2008) for the CPT and SPT data, respectively. Post liquefaction settlement of unimproved ground is estimated to be in the order of 300 to 600 mm based on the information provided. As the bridge will be founded in the underlying competent soils this settlement would not directly affect the bridge. However, the settlement would affect the connections between the bridge and the grade supported facilities (eg. parkade, embankment). Our interpretation of the standard penetration test (SPT) and CPT data in proximity to the proposed foundation location indicates that the fill would be expected to liquefy below the water table between about El. +1 to -4 m. The fill below El. -4 m has the potential to liquefy in discrete zones, and would be expected to exhibit cyclic mobility as the calculated factor of safety (FoS) against liquefaction was generally below 1.2. 5.3 Lateral Displacement Assessment Lateral displacement was assessed using the procedure presented in Bray and Travasarou (2007). Slope stability calculations were carried out using the program Slope/W (v.7.17). The calculated yield coefficient of the existing embankment geometry is 0.06 g. The model geometry, material parameters, and calculated failure surface are shown in Figure 4. The yield coefficient was calculated using post-liquefaction strength parameters. The estimated lateral displacement for the calculated yield coefficient is in the order of 400 to 500 mm.


The lateral displacement assessment only considered movement of the existing embankment fill towards the railway track. Lateral displacement towards the river is considered to have been addressed by others for the waterfront park development. The design level seismic event for the densification was indicated to be 1:2475 for upland areas, and 1:475 for offshore areas in Document 2. Global and internal static stability of the existing embankment wall is assumed to have been addressed by others. 6. PILE DESIGN 6.1 General We understand that the foundation piles are to consist of seven, partially concrete filled, 914 x 15.9 mm steel pipe piles. Pile loads at the underside of pile cap were provided by AE. As such, the self weight of the pile is not considered in the loads. The factored static axial loads are in the order of 1750 kN (compression) and 575 kN (tension). The factored seismic axial loads are in the order of 1850 kN (compression) and 1160 kN (tension). The factored kinematic displacement axial loads are in the order of 2800 kN (compression) and 2175 kN (tension). The governing load condition is the kinematic displacement. The compression loads are highest in the landward piles, the tension loads are highest in the riverward piles. 6.2 Axial Resistance Based on the loads provided, it is recommended that the landward and riverward piles be driven at least 8 and 12 m, respectively, into the dense till-like / hard silt soils. Available subsurface information would indicate that this corresponds to a pile toe at (or below) El. -16 and -20 m, respectively. Under service loads pile settlement is expected to be small (i.e. < 20 mm) and elastic (i.e. deformation occurs rapidly after the addition of load). The axial resistance was estimated using the Beta Method () as outlined in Canadian Foundation Engineering Manual (CFEM) IV. The Beta parameters used were back from high strain dynamic testing (HSDT) performed during the construction of the waterfront park. The HSDT test results were provided by Worley Parsons. Our pile embedment recommendations are based on the assumption HSDT will be performed on at least one each of the landward and riverward piles. This allows for an improved geotechnical resistance factor compared to empirical design without confirmatory testing (0.5 vs. 0.4 for static condition). For compression resistance a geotechnical resistance factor of 0.6 has been used for kinematic conditions. For tension resistance, a geotechnical resistance factor of 0.3 has been used for static conditions and 0.4 for kinematic conditions.


6.3 Equivalent Spring Constant Our analyses were performed using the computer program LPILE Plus 5.0. Tables 2 and 3 (attached) present the estimated equivalent soil spring value (Kequiv) for a single 914 mm diameter steel pipe pile. It should be noted that the analyses only considered resistance from soil below the prevailing site grade at the base of the wall (El. +4 m) and that the MSE embankment fill is ignored. The Kequiv values for the liquefied soils are estimated using the ‘p-multiplier’ procedures outlined in PEER 11/04. Under seismic inertial loading conditions, the lateral pile design should be carried out using both non-liquefied and liquefied Kequiv values. The Kequiv values have been estimated for lateral displacements of 2, 5, 10, 20, 30, 50, 75 and 100 mm. Below 22 m, the Kequiv values may be taken as equal to the calculated spring constants at 22 m depth. Since the pile layout is asymmetric, the Kequiv values presented in Tables 2 and 3 must be modified by group effect p-multipliers to account for the different longitudinal and transverse layout of the piles. Sketch 1 provides group effect p-multiplier values for the longitudinal and transverse direction. The group effect p-multiplier values would apply for all depths when using the non-liquefied soil spring values. However, when using the liquefied soil spring values the p-multiplier values would only apply below 12 m depth (i.e. p-multiplier equal to 1.0 above 12 m). 7. CONSTRUCTION AND RELATED ISSUES 7.1 Damage to the MSE Wall System We envisage that installation of the four landward piles will likely displace/damage the existing wall facing and reinforcement. The extent of displacement/damage is difficult to predict. It should be noted that, due to the short geogrid length, reconstructing the same wall system would not be practical in front of the piles. A concrete facing wall may be a possibility. The installation of the three riverward piles away from the wall face (i.e. > 4 m) is not be expected to affect or damage the installed geogrids to the same extent, provided the installed length is as per Document 3. However, pile installation may affect the EPS as noted below. We are aware that there have been previous issues associated with slumping of the upper portion of the MSE wall in the vicinity of the Basketball Courts over a distance of about 5 m. The mechanism responsible for the slumping is not known, but repairs were apparently performed early last year.


7.2 Location and Extents of EPS Document 5 shows the centre of the riverward piles is at a distance of about 6.2 m from the toe of the MSE retaining wall. However, referencing Document 4, this would indicate that the centre of pile could be just outside (0.3 m) or well within (1.2 m) the EPS zone. The discrepancy in EPS location between Documents 4 and 5 must be resolved prior to pile installation. We understand this will be possible during excavation to the underside of pile cap as it is shown as being 1.8 m thick. We expect that it would be difficult to accurately position and install a pile fully, or partially, on EPS (or through EPS) due to the low crushing strength of the EPS relative to the load of the piling leads. This should be reviewed with the piling contractor, who would likely prefer a fully granular working surface to install from. 7.3 Pile Driving In order to achieve the minimum required embedment depth into dense native soils, the 914 x 15.9 mm steel pipe piles are to be installed open ended. If driving shoes are used, they must be flush with the outer diameter of the pile. The piling contractor should be made aware that driving piles through the dense to very dense native till-like soils may be difficult. The construction methodology and equipment used should be capable of installing the piles to the minimum required embedment depth. If necessary, pile installation may require pre-drilling, churning or similar to achieve the minimum required embedment depth. A preliminary assessment indicates that a driving energy in the order of 200 kJ will be required to develop adequate resistance, which is equivalent to an APE D62-42 diesel hammer. TEL should be engaged to develop termination criteria for the piles and to provide full time inspection of the pile installation. Pile termination criteria should be determined by a wave equation analysis, or similar, which would require information on the actual pile driving equipment to be used. As noted above, HSDT of at least one each of the landward and riverward piles should be carried out in order to confirm the design assumptions and justify the geotechnical resistance factors used to estimate the vertical compression resistance. TEL can perform this testing and analysis, and would be pleased to provide this service if required.


References Bray and Travasarou; “Simplified Procedure for Estimating Earthquake-Induced Deviatoric Slope Displacements”; Journal of Geotechnical and Geoenvironmental Engineering, April 2007 Idriss and Boulanger; “Soil Liquefaction During Earthquakes”; Earthquake Engineering Research Institute Monograph 12 (MNO-12); 2008 Pacific Earthquake Engineering Research Centre; “Recommended Design Practice for Pile Foundations in Laterally Spreading Ground”; June 2011 Youd, et al.; “Liquefaction Resistance of Soils: Summary of Report from the 1996 NCEER and 1998 NCEER/NSF Workshops on Evaluation of Liquefaction Resistance of Soils”; Journal of Geotechnical and Geoenvironmental Engineering, October 2001 Zhang, et al.; “Estimating liquefaction-induced ground settlements from CPT for level ground”; Canadian Geotechnical Journal, September 2002

104292 Thurber Statement:75443-52 Thurber Letterhead 6/15/11 9:21 AM Page 1

SLC20110614

104292 Thurber Statement:75443-52 Thurber Letterhead 6/15/11 9:21 AM Page 2

NEW WESTMINSTER, BC FIGURE 1AS SHOWNMAY 15, 2013

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1. SECTION TAKEN FROM TROW ASSOCIATES INC."GEOTECHNICAL FACTUAL REPORT AND SEISMICANALYSIS PRELIMINARY RECOMMENDATIONS AT30% DESIGN STAGE FOR PROJECT MASTER PLAN"REPORT DATED APRIL 9, 2010.

NEW WESTMINSTERWATERFRONT INVESTIGATION

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FIGURE 3TEL 17-123-786 - 4th Street Pedestrian Bridge, New Westminster, B.C.

Surface Response Spectra (1:475) vs.CAN/CSA-S6-06 Response Spectra (1:475)

EERA Output

CAN/CSA-S6-06

CAN/CSA-S6-06 (Clause 4.4.7.3)

CAN/CSA-S6-06PHA = 0.267 (from NBC 2010 seismic hazard website)Soil Class = 3Importance Facotor = 1.01.0 Csm

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6/12/2013 S:\Data\Projects\17\123 Associated Eng\786\MCB\Seismic\17‐123‐786 SSRA Summary of Results

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Delta Kequiv Delta Kequiv Delta Kequiv Delta Kequiv Delta Kequiv Delta Kequiv Delta Kequiv Delta Kequiv Delta Kequiv Delta Kequiv Delta Kequiv

(mm) (MN/m) (mm) (MN/m) (mm) (MN/m) (mm) (MN/m) (mm) (MN/m) (mm) (MN/m) (mm) (MN/m) (mm) (MN/m) (mm) (MN/m) (mm) (MN/m) (mm) (MN/m)

2 20 2 40 2 60 2 34 2 42 2 50 2 58 2 90 2 100 2 110 2 1205 20 5 40 5 49 5 34 5 42 5 50 5 58 5 90 5 100 5 110 5 120

10 13 10 26 10 33 10 34 10 40 10 50 10 58 10 84 10 97 10 110 10 12020 9 20 17 20 23 20 30 20 32 20 39 20 47 20 66 20 77 20 88 20 9530 7 30 13 30 19 30 26 30 28 30 34 30 42 30 58 30 67 30 77 30 83

Depth 11 m

TEL 17-123-786Calculated Equivalent Spring Constant 'Kequiv' for 914 mm Diameter Steel Pipe Pile

Non-Liquefied Condition

Depth 1 m Depth 2 m Depth 3 m Depth 4 m Depth 5 m Depth 6 m Depth 7 m Depth 8 m Depth 9 m Depth 10 m

TABLE 2

S:\Data\Projects\17\123 Associated Eng\786\MCB\Piles\17-123-786 914 Static p-y -MCB 6/12/2013

50 6 50 12 50 18 50 24 50 26 50 33 50 40 50 55 50 64 50 73 50 7975 6 75 12 75 17 75 24 75 26 75 32 75 38 75 54 75 62 75 71 75 77

100 6 100 11 100 17 100 23 100 25 100 31 100 37 100 52 100 60 100 69 100 75



2 241 2 241 2 241 2 241 2 241 2 241 2 258 2 276 2 276 2 276 2 2765 120 5 120 5 120 5 120 5 120 5 120 5 129 5 138 5 138 5 138 5 138

10 71 10 71 10 71 10 71 10 71 10 71 10 76 10 81 10 81 10 81 10 8120 43 20 43 20 43 20 43 20 43 20 43 20 46 20 49 20 49 20 49 20 4930 35 30 35 30 35 30 35 30 35 30 35 30 37 30 40 30 40 30 40 30 4050 21 50 21 50 21 50 21 50 21 50 21 50 23 50 24 50 24 50 24 50 2475 16 75 16 75 16 75 16 75 16 75 16 75 18 75 19 75 19 75 19 75 19

13 13 13 13 13 13 14 15 15 15 15

Depth 12 m Depth 13 m Depth 14 m Depth 15 m Depth 16 m Depth 17 m Depth 18 m Depth 19 m Depth 20 m Depth 21 m Depth 22 m

100 13 100 13 100 13 100 13 100 13 100 13 100 14 100 15 100 15 100 15 100 15

S:\Data\Projects\17\123 Associated Eng\786\MCB\Piles\17-123-786 914 Static p-y -MCB 6/12/2013



2 20 2 40 2 21 2 3 2 3 2 4 2 5 2 13 2 14 2 15 2 175 20 5 40 5 21 5 3 5 3 5 4 5 5 5 13 5 14 5 15 5 17

10 13 10 26 10 14 10 3 10 3 10 4 10 5 10 12 10 14 10 15 10 1720 9 20 17 20 10 20 2 20 3 20 3 20 4 20 9 20 11 20 12 20 1330 7 30 13 30 8 30 2 30 2 30 3 30 3 30 8 30 9 30 11 30 12

Depth 11 m

TEL 17-123-786Calculated Equivalent Spring Constant 'Kequiv' for 914 mm Diameter Steel Pipe Pile

Liquefied Condition

Depth 1 m Depth 2 m Depth 3 m *1 Depth 4 m Depth 5 m Depth 6 m Depth 7 m Depth 8 m Depth 9 m Depth 10 m

TABLE 3

S:\Data\Projects\17\123 Associated Eng\786\MCB\Piles\17-123-786 914 Seismic p-y -mcb 6/12/2013

50 6 50 12 50 7 50 2 50 2 50 3 50 3 50 8 50 9 50 10 50 1175 6 75 12 75 7 75 2 75 2 75 3 75 3 75 7 75 9 75 10 75 11

100 6 100 11 100 7 100 2 100 2 100 2 100 3 100 7 100 8 100 10 100 10



2 241 2 241 2 241 2 241 2 241 2 241 2 258 2 276 2 276 2 276 2 2765 120 5 120 5 120 5 120 5 120 5 120 5 129 5 138 5 138 5 138 5 138

10 71 10 71 10 71 10 71 10 71 10 71 10 76 10 81 10 81 10 81 10 8120 43 20 43 20 43 20 43 20 43 20 43 20 46 20 49 20 49 20 49 20 4930 35 30 35 30 35 30 35 30 35 30 35 30 37 30 40 30 40 30 40 30 4050 21 50 21 50 21 50 21 50 21 50 21 50 23 50 24 50 24 50 24 50 2475 16 75 16 75 16 75 16 75 16 75 16 75 18 75 19 75 19 75 19 75 19

13 13 13 13 13 13 14 15 15 15 15

Depth 12 m Depth 13 m Depth 14 m Depth 15 m Depth 16 m Depth 17 m Depth 18 m Depth 19 m Depth 20 m Depth 21 m Depth 22 m

100 13 100 13 100 13 100 13 100 13 100 13 100 14 100 15 100 15 100 15 100 15

Note:*1 Kequiv reduced to account for interaction with underlying liquefied zone

S:\Data\Projects\17\123 Associated Eng\786\MCB\Piles\17-123-786 914 Seismic p-y -mcb 6/12/2013

REVISIONS

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PEDESTRIAN OVERPASS

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KARAM BAHI

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PEDESTRIAN OVERPASS

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GENERAL ARRANGEMENT

KARAM BAHI

A4

2297-104

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June 17, 2013

REFERENCEONLY