guidelines for design of dams for earthquake

8/11/2019 Guidelines for Design of Dams for Earthquake

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guidelines for

design of dams

for earthquake

- lu#_,

AUSTRALIAN NATIONA L

COM MITTEE ON

LAR GE DAMS


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GUIDELINES FOR

D E S IG N O F D A M S

FOR EARTHQUAKE

AUG UST 1998

AUSTRA LIAN NATIONA L

COM MITTEE ON

LARGE DAM S


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A U S T R A L IA N N A T IO N A L C O M M IT T E E O N L A R G E D A M S

G U ID E L IN E S FO R D E S IG N O F

D A M S FO R E A R T H Q U A K E

AUGUST 1998

IM P O R T A N T D IS C L A IM E R

"ANCOLD and its Members, and the Convenor, Members and Assistants of the

Working Group which developed these Guidelines do not accept responsibility

for the consequences of any action taken or omitted to be taken by any person,

whether a purchaser of this publication or not, as a consequence of anything

contained in or omitted from this publication. No persons should act on the basis

of anything contained in this publication without taking appropriate professional

advice in relation to the particular circumstances".

ANCOLD Guidelines for Design of Dams for Earthquake


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T A B LE O F C O N T E N T S

Page No.

F O RE W O RD

ANCO LD WORKING GRO UP MEMBER SHIP L IST

NTRODUCTON1

EARTHQUAKES AND THEIR CHARACTERISTICS2

2.1Earthquake Mechansms and Termnoogy2

22EarthquakeGoundMoon2

23SuaceRupue3

24MagntudeandInensty3

2.5Changes to Seismic Waves Near the Ground Surface4

2.6Attenuation and Amplification of Ground Motion4

27Reservor InducedSesmcty5

EARTHQUAKE HAZARD IN AUSTRALIA 6

31Gnea6

32Mechansmo Earhquakes8

33EahquakeDphs9

34Evauaono Sesmc Hazad10

35Aenuaon11

3.6MaximumCredbe Earthquake Magntude11

3.7Estimates of Ground Motion and Response Spectra at a Site12

38EarthquakeHazardMaps12

SELECTION OF DESIGN EARTHQUAKE 16

41Denons16

42Seection of the Design Earthquake20

4.3 Selection of the Operating Basis Earthquake (OBE) 32

44Concurrent Load Combinations32

45Earthquakes Induced by the Reservor33

46Response Spectra and Acceerograms *33

D E S I GN OF E M BA N K M E N T D A M S A N D A N A LY S IS OF

LQUEFACTON33

5.1 Effect of Earthquake on Embankment Dams3 3

5.2 General ("Defensive") Design Principles for Embankment Dams 34

5.3Liquefaction of Dam Embankments and Foundations36

ANCOLD Guidelines for Design ofDams for Earthquake


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6. SEISMIC STABILITY ANALYSIS OF EMBANKMENTS 57

61Peame57

62Pseudo-Sac Anayss57

6.3 Simplified Methods of Deformation Analysis59

6.4Post Liquefaction Stability and Deformation Analysis63

65Numerca Mehods65

66PoposedGudenes67

7.ANALYSIS AND DESIGN OF CONCRETE DAMS69

7.1 Past Performance of Concrete Dams in Earthquakes69

72DeensveDesgnMeasues70

73AnayssMhods71

7.4Design Earthquake and Hydrodynamic Loads82

75DsgnCea83

76Dynamc Maea Popetes86

8APPURTENANT STRUCTURES87

81noducon87

82PerormanceRequremens87

83nakeTows89

RE F E RE NC E S

APPEND IX A

TERMS OF REFERENCE

APPENDIX B

TYPICAL EASTERN AUSTRALIAN PEAK

G RO UND A C C E L E RA T IO N V S A E P

RE S PO NS E S PE C TR UM FO R 1 in 1000 A E P

MODIFIED MERCALLI SCALE

APPENDIX C

EXTRACTS FROM CANADIAN DAM SAFETY

GUIDEL INES

APPENDIX D

ADDITIONAL INFORMATION ON ACCEPTABLE

RISKS



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F O R E W O R D

Even in the matter of earthquakes, Australia can be considered the "Lucky Country" in not being on the

edge of major tectonic plates. Neighbouring countries like New Zealand and Indonesia are renowned for

their volcanoes and frequent earthquakes. Australia is relatively earthquake free by comparison and

earthquakes were seldom considered in early dam designs.

Certainly there were some zones of known activity such as the Adelaide Hills and the Western

Australian wheat belt, and whilst major damage had occurred, it was not on the same scale as in other

countries.

In 1979, the Standards Association of Australia produced the "Earthquake Code" AS2121. It showed

zones of seismic activity and recommended methods of determining loads on building structures. The

development of this code was based largely on statistics of historic earthquakes, for which there were

relatively short term records.

However, several major earthquakes subsequently occurred in areas indicated by the code as having

negligible earthquake risk, the most notable being the 1989 earthquake at Newcastle (Magnitude 5.6) in

which 12 people died and the Tennant Creek Earthquake in 1988 (Magnitude 6.8). This led to the

introduction of a new earthquake code (AS 11 70.4-1993)which included data from m ore widespread and

reliable seismographs and furthermore considered the all important geological situations.

In parallel with these developments, analytical methods used by dam engineers were improving beyond

the simplistic application of a horizontal force equating to seismic acceleration. Improvements were

based on the observed fact that earth dams subjected to earthquakes had slumped vertically rather than

fail by slipping of a face as indicated by the simplistic analyses.

Methods of analysing slumping were developed, and further supplemented by sophisticated finite

element analyses which, by utilising modem computer power, give an ability to undertake rigorous

analyses of dams where necessary.

This ANCOLD Guideline brings together improved appraisals of the earthquake loadings that a dam

may suffer and then describes appropriate methods for analysis and evaluation. Whilst specific to the

Australian considerations, the majority of this guideline could be applied to dam structures throughout

the world. The IC O LD Bulletins No. 46(198 3), No. 52(198 6), No. 62(198 8) and No. 72(1989 ) are

parallel documents in this regard, although not including recent advances.

This guideline is a major contribution to dam engineering and the voluntary work by the ANCOLD

subcommittee has unselfishly provided their experience to the dam building community and indeed the

wider community. Our appreciation goes to Prof Fell and his team for producing this valuable guideline.

This guideline is not a design code, and dam designers must continue to apply their own considerations,

judgements and professional skills when designing dams to resist earthquakes. As time goes on there

will no doubt be improved data and design tools to help the designer and it is intended that this guideline

will be updated as circumstances dictate. ANCOLD welcomes contributions to discussion on this

guideline which will assist with future revisions.

/

JOHN PHILLIPS

Chairman, ANCOLD



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MEMBERSHIP OF THE ANCOLD WORKING GROUP FOR

G U ID E L IN E S F O R T H E D E S IG N O F D A M S F O R E A R T H Q U A K E

4

Robin Fell

School of Civil Engineering, University of New South Wales

Gamini Adikari

Snowy Mountains Engineering Corporation, Victoria

John Bosler

Snowy Mountains Engineering Corporation, Cooma, New South Wales

Brian Cooper

Dam s and C ivil Section, Public W orks and Services Department, NS W

Peter Foster

Works Consultancy Services, Power Engineering, New Zealand

Gary Gibson

Seismology Research Centre, RMIT, Melbourne

Sergio Giudici

Hydro-Electric Commission, Hobart

Nasser Khalili

School of Civil Engineering, University of New South Wales

Ian Landon-Jones

Dams S afety Group, Sydney Water, New S outh Wales

Kevin McCue

Australian Seismological Centre, Canberra

Len McDonald

Dams and Civil Section, Public Works and Services Department, NSW

Brian Shannon

Water Resources, Department of Primary Industries, Queensland

David Stapledon

Geotechnical Consultant, Adelaide, South Australia

John Waters

Geo-Eng Pty Ltd, Perth, Western Australia

Ron Wyburn

Halcrow Water Power, Victoria



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1. INTRODUCTION

Public awareness of the potential for damage

and loss of life in Australia from earthquakes

was highlighted by the Newcastle earthquake in

December 1989. This was a Magnitude 5.6

(M5.6) event, and because of its proximity to

Newcastle, and local ground conditions, caused

approximately $1 billion damage.

Dam engineers in Australia have been

conscious of earthquakes for many years, but it

was the earthquakes at Tennant Creek in 1988,

which were M6.3, M6.4, M6.7, with a total fault

scarp length of 32km which raised the question

most acutely as to whether dams in Australia

could be subject to large earthquakes, and if so,

could they withstand them without resultant

loss of the facilities and lives, property, and

environmental values downstream. Other large

earthquakes in the M6 to M7 range had

occurred in Australia, the most notable being in

Meckering in 1968 (M6.9), but the Tennant

Creek event was critical because it occurred in

an area which had previously been regarded as

virtually free of earthquakes.

Recent assessments of earthquake ground

motions for some large Australian dams have

been based on the assumption that the

maximum credible earthquake is M7.5, which is

large by any standards. The seismologists

involved in these studies indicate that on the

available evidence, such earthquakes, ie. M7.5,

could occur anywhere in Australia.

In general, it is not possible to identify active

faults which might cause such earthquakes. For

example, there had been no movement on the

Tennant Creek fault for more than 200,000

years (Crone and Machette, 1992), so the

question arises, can it occur at, or close to any

damsite? Peak ground accelerations close to a

M 7.5 earthquake can be very high.

The past performance of dams in earthquake

h&s been very good, with few dams suffering

major damage. Where this has occurred, it has

been due to liquefaction in the dam or the

foundation. Very few of these dams have

breached and released a flood wave. However,

several might have breached, if the reservoir

level had been higher at the time of the

earthquake. Seed (1979), USCOLD (1992),

ICOLD (1986), NSWDSC (1993) and Hinks

and Gosschalk(1993) give some details.

The approach taken by seismologists in

Australia is to use statistical analysis to predict

the frequencies of recurrence of ground

motions. This typically results in 1 in 1000

AEP peak ground accelerations of 0.15g, 1 in

10,000 AEP s0.35g, and 1 in 100,000 AEP

~0.5g. These are large loadings and it is likely

that assessment of many of the existing dam s in

Australia for such loads could indicate some

deficiencies to either the dam or appurtenant

structures. New dams would also need (less)

expensive additional design features to cope

with earthquake.

In recognition of the need to provide some

guidance to dam engineers and owners in

Australia, ANCOLD established a Working

Group to prepare Guidelines for the Design of

Dams for Earthquake. The Working Group was

established in September 1993, and took over

from an earlier ANCOLD Working Group

preparing Guidelines on Seismic Analysis and

Design of Embankment Dams. The Terms of

Reference for the Working Group are in

Appendix A.

These guidelines are to cover all types of dams,

including tailings dams, and apply to existing

and new dams. They cover the selection of the

design earthquake, analysis and design of

embankment and concrete dams, and

appurtenant structures.

The guidelines are not meant to be used as a

design code, and of necessity, do not include

complete details of all the analysis and design

methods which are recommended. The area is

rapidly evolving, and those involved in the

analysis and design of dams for earthquake

should refer to the references given, and to

more recent publications so as to be fully

informed. In some situations it will be

necessary to seek specialist advice.


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2. EARTHQUAKES AND

T H EIR CH A R A CT ER IS T ICS

2.1 Earthquake Mechanisms and

Terminology

An earthquake is the motion that is produced

when stress within the earth builds up over a

long period of time until it eventually exceeds

the strength of the rock, which then fails and a

break along a fault is produced. It may take

tens, hundreds or thousands of years for the

stress to build up in a particular area, and it is

then released in a few seconds. Part of the

energy is transmitted away as seismic waves

and part of it as heat.

The fault displacement in a particular

earthquake may vary from centimetres up to a

few metres in a great earthquake. Once

ruptured, the fault is a weakness which is more

likely to fail in future earthquakes, so a large

total displacement may build up from many

earthquakes over a long period of time. This

may eventually measure kilometres for thrust

faults produced by compression, or hundreds of

kilometres for horizontal strike-slip faults such

as the San Andreas.

The point on the fault surface where a

displacement commences is called the

hypocentre or focus, and the earthquake

epicentre is the point on the ground surface

vertically above the hypocentre. The

displacement usually propagates along the fault

in one direction from the hypocentre, but

sometimes it propagates in both directions.

Energy release is near but not exactly at the

hypocentre.

The hypocentral distance from an earthquake to

a point is the three dimensional slant distance

from the hypocentre to the point, while the

epicentral distance is the horizontal distance

from the epicentre to the point.

2.2 Earthquake Ground Motion

Earthquake ground vibration is recorded by a

seismograph or a seismogram. Most modem

seismographs record three components ol

motion: east-west, north-south and vertical.

The rupture time for small earthquakes is a

fraction of a second, for earthquakes of

magnitude 5.0 it is about a second, and for large

earthquakes may be up to tens of seconds.

However the radiated seismic waves travel al

different velocities, and are reflected and

refracted over many travel paths, so the total

duration of vibrations at a site persists longei

than the rupture time, and shows an exponentia

decay.

Several types of seismic wave are radiated from

an earthquake. Body waves travel in three

dimensions through the earth, while surface

waves travel over the two dimensional surface

like ripples on a pond. There are two types oi

body wave (P and S waves), and two types oi

surface wave (Rayleigh and Love waves).

Primary or P waves are ordinary sound waves

travelling through the earth. They are

compressional waves with particle motion

parallel to the direction of propagation.

Secondary or S waves are shear waves, with

particle motion at right angles to the direction of

propagation. The amplitude of S waves from an

earthquake is usually larger than that of the P

waves.

P waves travel through rock faster than S

waves, so they always arrive at a seismograph

before the S wave.

The frequency content of earthquake ground

motion covers a wide range of frequencies up to

a few tens of hertz (cycles per second). Most

engineering studies consider motion between

about 0.2 and 25 Hz.

The amplitude, duration and frequency content

of earthquake ground motion at a site depend on

many factors, including the magnitude of the

earthquake, the distance from the earthquake to

the site, and local site conditions.

The larger the earthquake magnitude, the

greater the amplitude (by definition a factor of

ten for each magnitude unit), the longer the

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duration of motion, and the greater the

proportion of seismic energy at lower

frequencies. A small earthquake has low

amplitude (unless it is very close), short

duration, and has only high frequencies.

The smaller the distance from an earthquake to

the site, the higher the amplitude. The duration

is not strongly affected by distance. High

frequencies are attenuated by absorption within

the ground more quickly than low frequencies,

so at greater distances the proportion of seismic

vibration energy at high frequencies will

decrease.

2.3 Surface Rupture

Surface rupture is a relatively rare phenomenon

which occurs when a fault break reaches the

ground surface. It may produce a vertical or

horizontal offset (or both) with a displacement

of millimetres to a few metres, and a length

from metres to tens of kilometres.

Because rock near the surface is relatively

weak, few earthquake hypocentres occur in the

top one or two kilometres. It is common for

surface sedimentary rocks to be folded in

response to faulting at depth, giving a

monocline and scarp at the surface, but without

a surface fault.

Most earthquakes, especially most larger

earthquakes, occur on existing faults. This is

because faults are weaker than surrounding

unbroken rock, and are much more likely to fail

again when stress rebuilds.

A site will have surface rupture potential if an

existing fault is found which has been active in

the recent geological past (perhaps the past few

million years). This will be quite rare, and

possibly be difficult to establish. It will usually

be easier to show that a site with simple surface

geology has no faulting history, than to show

that a site with complex geology has suffered

recent faulting.

2.4 Magnitude and Intensity

Earthquakes vary enormously in size. In 1935

Richter defined a magnitude scale to indicate

the size of an earthquake. For the Richter local

magnitude scale, ML, the logarithm of the peak

ground displacement is taken and an empirical

correction depending on the distance from

earthquake to seismograph is subtracted. The

resulting values are averaged for all the

seismographs that have recorded the

earthquake.

Other magnitude scales have been defined,

including moment magnitude, and while not

exactly the same as the R ichter local magnitudes,

they give similar values that can range from 0.0

to over 9.0. For each unit of magnitude there is

a tenfold increase in ground displacement, and a

thirtyfold increase in seismic energy release.

Another measure of earthquake size is the fault

area, or the area of the fault surface which is

ruptured. The fault area ruptured in an

earthquake depends on the ma gnitude and stress

drop in the earthquake. For a given magnitude,

a higher stress drop will give a smaller rupture

area. Typically, a magnitude 4.0 earthquake

ruptures a fault area of about 1 square

kilometre, magnitude 5.0 about 10 square

kilometres, and magnitude 6.0 about 100 square

kilometres (perhaps 10 by 10 kilometres).

Earthquake Intensity is a measure of the effect

of the seismic waves at the surface, and is

normally given on the Modified Mercalli

Intensity scale, a copy of which is attached in

Appendix B. This is an arbitrary scale defined

by the effects observed (whether sleeping

people were woken, trees shaken, etc) and on

the amount of damage caused. Normally the

maximum intensity occurs near the epicentre of

the earthquake, and intensity then decreases

with distance. However, this may be affected

by the orientation of the earthquake rupture, or

by local ground conditions such as topography

or surface sediments.

The earthquake recurrence or seismicity

(seismic activity) of an area must take the range

of earthquake sizes into account. There are

many more small earthquakes than large. In


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most places around the earth there are about ten

times as many earthquakes exceeding

magnitude 3.0 than there are exceeding

magnitude 4.0, and ten times as many again

exceeding magnitude 2.0. In seismicity studies,

the logarithm of this factor is called the b value,

so a value of 1.0 is typical. The b value may be

1.3 or higher if there are many small

earthquakes, or 0.7 or lower if there are few

small earthquakes.

2.5 Changes to Seismic Waves Near

the Ground Surface

The energy in seismic waves depends upon

their amplitude and the physical properties of

the material through which they are passing.

When waves pass from high stiffness material

(eg. rock at depth) into lower stiffness material

(eg. near-surface rock, or sediments) they are

reflected towards the vertical and their

amplitude increases. Their amplitude also

increases as they approach the earth's (free)

surface, at which they are reflected. The nature

and extent of free surface amplification varies

with topography, even in fresh, strong rock.

Changes in soil thickness above an irregular

bedrock surface can give complex surface

amplification that varies with earthquake wave

duration.

Resonance in the surface sediments causes

amplification at particular frequencies,

especially at the natural frequency of the

sediments. This depends on the thickness and

elastic properties of the sediments. Earthquake

motion recorded on hard rock includes all

frequencies up to a value that depends on

magnitude, while that recorded on soft

sediments is usually dominated by the resonant

frequency.

In surface sediments, high frequency vibrations

are attenuated much more with distance than

low frequencies. If sediments are very thick,

much of the high frequency motion will be lost

and peak surface accelerations will be low, even

if resonance has amplified motion at the low

resonant frequency.

2.6 Attenuation and Amplification of

Ground Motion

Earthquake ground motion attenuates with

increasing distance from the source due to

radiation and hysteretic damping. High

frequency motion is attenuated more quickly

with distance than lower frequency motion.

For estimates of peak ground acceleration,

attenuation is allowed for by using an

attenuation function of the form

a =b,eb2Mr*3

where a =acceeration

R=foca dstance

M=Magntude

b jb are constants, which

vary considerably over the

world.

Some earthquake hazard studies use the Esteva

and Rosenblueth (1969) attenuation functions,

which give peak ground velocity (mm/s), peak

ground acceleration (mm/s2) and Modified

Mercalli Intensity (IMM) at an epicentral distance

x kilometres from an earthquake at depth z

kilometres with local magnitude M. The

equations are:

R

=

Vx2 +z2 +400

Vpeak

= 160 e10 M R"17

Speak

=

20000 e0 8 mR20

Im m

=

loge(2980e15MR-23)

Because of the 400 term in the expression for R,

corresponding to a minimum R of 20

kilometres, these equations give low values of

ground motion at distances closer than a few

kilometres.

These relations were determined using

Califomian data, and should only be used with

magnitudes determined using a compatible

function. If the magnitudes computed for

seismographs at different distances vary, then

the attenuation function is invalid for the area .

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In selecting attenuation relationships care is

needed, and attention paid to the mechanism of

the source earthquake, eg. whether shallow

intraplate, or deep crustal boundary

earthquakes.

Weak surface materials absorb seismic energy

rather than transmit it unchanged, thus tending

to reduce amplitudes at the surface. The

amount of attenuation depends on the properties

of the materials, and especially on their

thickness.

Near-surface layers will vibrate preferentially at

their own natural frequencies, depending on

their thickness and elastic properties. The

earthquake motion at the natural frequencies of

the near-surface layers is amplified, while

motion at other frequencies may be little

affected or even attenuated. The amplification

effect can be especially pronounced for deep

soft sediments such as those underlying M exico

City, but in deep, stiff sediments subject to high

frequency earthquake, attenuation may result.

Dams (like all other structures) have natural

frequencies of their own depending on their

mass and stiffness, usually in the range from

about 0.5 hertz to about 5 hertz for embankment

dams and 2 hertz to 20 hertz for concrete

gravity dams.

2.7 Reservoir Induced Seismicity

Reservoirs may induce seismicity by two

mechanisms. Either the weight of the water

may change the stress field under the reservoir,

or the increased ground water pore pressure

may decrease the stress required to cause an

earthquake. In either case, reservoir induced

seismicity (RIS) will only occur if relatively

high stresses already exist in the area. If the

stress has been relieved by a recent large

earthquake, say in the last few hundred years

for low seismicity areas like Australia, then RIS

is unlikely to occur.

/

RIS events initially usually occur at shallow

depth under or immediately alongside a

reservoir. As years pass after first filling, and

groundwater pore pressure increases permeate

to greater depths and distances, the events may

occur further from the reservoir. This occurs at

a rate of something like one kilometre per year.

RIS is experienced under new reservoirs,

usually starting within a few months or years of

commencement of filling, and usually not

lasting for more than about twenty years. Once

the stress field and the pore pressure fields

under a reservoir have stabilised, then the

probability of future earthquakes reverts to a

value similar to that which would have existed

if the reservoir had not been built. Most of the

earthquake energy does not come from the

reservoir, but from normal tectonic processes.

The reservoir simply acts as a trigger.

In areas with horizontal tectonic compression

and reverse faulting, like Australia, filling a

reservoir should increase the vertical minimum

principal stress and reduce the chance of an

earthquake under the reservoir. This has been

called reservoir induced a seismicity. However,

in some cases earthquakes could then be

induced by later releasing water from the

reservoir. Alternatively the change in stress

during filling could induce earthquakes beside

the reservoir rather than under it, although this

stress change is less pronounced.

It has been suggested that filling a reservoir will

cause compression under it, increasing the pore

pressure of the existing groundwater, and so

tend to induce earthquakes even in areas of

horizontal compression. Stress change induced

seismicity, either direct or through this indirect

mechanism, should occur soon after filling. It

may then cause seasonal variations in

seismicity, sometimes lagging a few weeks or

months behind water level.

Pore pressure induced seismicity is normally

delayed, and may occur years after filling. Pore

pressure increases always tend to induce events.

If there is a major fault near the reservoir, RIS

can produce earthquakes exceeding magnitude

6.0 (Xinfengjiang, China, 1962, M6.1; Koyna,

India, 1967, M6.3). Such events will only occur

if the fault is already under high stress. A

number of Australian reservoirs have triggered

earthquakes exceeding magnitude 5.0

(Eucumbene, 1959, M5.0; Warragamba, 1973,

M5.0; Thomson, 1996, M5.2).


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MPP*8-

It is more common for a reservoir to trigger a

large number of small shallow earthquakes,

especially if the underlying rock consists of

jointed crystalline rock like granite (Talbingo,

1973 to 1975; Thomson, 1986 to 1995). These

events possibly occur on joints rather than

established faults, so are limited in size, and

only give magnitudes up to 3 or 4. There is no

hazard from such low magnitude reservoir

induced earthquakes, even if they occur

regularly. Their shallow depth means that they

may often be felt or heard.

RIS has been observed for over one hundred

reservoirs throughout the world, and small

shallow induced events have probably occurred

under many others. A relatively high

proportion of reservoirs with RIS seismograph

networks do record such activity. A high

proportion of RIS examples occur in intraplate

areas, with above average rates in China,

Australia, Africa and India.

It is not easy to predict whether a reservoir will

experience RIS because the stress and strength

at earthquake depths are not easily measured.

For the same reason, prediction of normal

tectonic earthquakes has been unsuccessful in

most parts of the world.

It seems that RIS with many small events is

more likely to occur in intraplate areas with

near surface crystalline rocks like granite, rather

than sedimentary rocks. A larger magnitude

RIS event can only occur if there is an existing

fault of sufficient dimension that is late in its

earthquake cycle (the stress is already

approaching the strength of the fault).

3. EARTHQUAKE HAZARD

IN AUSTRALIA

The understanding of the hazard imposed by

earthquakes in Australia is critical to selection

and application of design earthquakes for dams.

Hence, a relatively detailed discussion on the

topic follows. This is largely taken from Gibson

(1994).

3.1 General

The Australian continent is within a tector

plate shared with Southern India, so all of

earthquakes are intraplate. The pla

boundaries to the north and east are among t

most active on the earth. Possibly as a result

this, Australia is one of the most actr

intraplate areas on the earth. Despite this, tf

hazard is quite low when compared with acth

interplate areas.

Most people in Australia can expect to feel

earthquake about every five or ten year

although many of these may not be recognisf

as an earthquake. Most Australian earthquake

that are reported are heard with a noise like

distant quarry blast or thunder, with possibly

slight vibration being felt.

Only a proportion of earthquakes that are fel

perhaps about one in twenty, will cause som

damage in their epicentral area. If they occur i

an inhabited area, most earthquakes larger tha

about magnitude 4.0 will cause some damage.

By contrast, in an active interplate area lik

New Britain or Bougainville in Papua Nev

Guinea, earthquakes are felt very often, c

average every week or two. These are normal

felt rather than heard, with any soun ds being thf

reaction of a building to the vibration rathe

than the earthqua ke itself.

PNC

havf

A very small proportion of these

earthquakes, perhaps about 1 in 500,

caused any damage in their epicentral area, an

o

aO

q.

*6

90o

. Q

If

s >

O do

a

o

*

8o

Q

< ?

o 00

o

o C O

o

o

a!?

H

Australian Earthquakes to 1994

Magntudes: *405o

Figure 1. Australian earthquakes with magnitude exceeding M L4.0 since 1850 (Gibson, 19 94).

Earthquakes on reverse faults usually give a

high stress drop, where the seismic energy

comes from a small source volume. High stress

drop earthquakes radiate a greater than average

proportion of their energy in higher frequencies.

Higher frequency motion implies higher

accelerations for a given energy release or

earthquake magnitude, but not necessarily

fewer cycles, than for similar magnitude

earthquakes in, say, West Coast USA.

Compression giving reverse and thrust faul

produces surface uplift. Therefore, earthquakf

are most likely to occur in areas whei

mountains are developing, and less likely und0)

jjquiiocScn

Pm+mnca\ Mo

Jogontt daw (

CtMHdol , A4

> Iw^BeNan

2030

*0

40

Figure 14. Relationship between stress ratios causing liquefaction and (N,),# values for clean sa

magnitude 7.5 earthquakes. (Seed and De Alba, 19 86).

2030

< N * 0

Figure 15. Relationship between stress ratios causing liquefaction and (N,)^ values for silty sa

magnitude 7.5 earthquakes (Seed and De Alba, 1986).

(iii) For earthquakes of magnitude other

than 7.5, correct the values of Tav/a'0 by

the factors in Table 13.

44



52/127

Table 13. Representative Number of Cycles and Corresponding Correction Factors (Seed and De

Alba, 1986).

Earthquake

Magnitude (M)

Number of

Representative Cycles

at 0.65 t

Factor to Correct

Abscissa of Curve in

Figures 14 and 15

8.5

26

0.89

7.5

15

1.00

6.75

10

1.13

6.0

5-6

1.32

5.25

2-3

1.5

It should be noted that the magnitude of

the earthquake has a significant

influence on whether liquefaction will

occur. It may be assumed that

liquefaction will not occur for

earthquakes of Magnitude 5 or less

(there are not sufficient cycles of

loading for small earthquakes).

However, where static liquefaction may

occur, even small earthquakes could

trigger failure. Morgenstem (1995)

discusses static liquefaction which is

likely to occur only in very

loose/poorly compacted, saturated soils

including dredged sand, mine

overburden dumps and mine tailings.

Salmon (1995) points out that when

earthquake loading is determined by a

probabilistic analysis of the history of

earthquakes in the seismotectonic zone

around the dam (as is done in

Australia), it is possible to determine

the contribution of different magnitude

earthquakes to the assessed ground

motion. Figure 16 gives an example

which shows that the major

contribution to the estimated ground

motion for a given PGA (in this case

the 1 in 1000 PG A) is from small

magnitude earthquakes near the dam.

Given the nonlinear relationship

implied in Tab le 13, and Figures 14 and

15, this has an important influence on

the assessment of the probability of

liquefaction, and should, if practicable,

be included in the assessment.

/


45


53/127

50

GO

OS

I t

o

10

50

Suniat9 0

M A G N IT U D E C O N T R IB U IIO N S

-t-

Sums5 6

lb.

34567

M A C N I IU O E ( M )

oisiANcr coNfRinunnN^

Nnle ConUfbolKjftS /ro


54/127

- 4

o

a.

c

o

2 3

- 0

O JomfoUowtki tf oL 09091

0 MurorrocM and KoboroiM UM 2)

A (ihihofo and K090 (Oil )

it Robvrlion (1902)

V MHehtll (1903)

O HoriNr > oL (1964)

ne '

bkul/lool

OO 002 003 01 0205 I

Mean Groin Sit*, O0 -mm

Figure 17. Variation of ratio with mean grain size (qc) measured in tsf 100 kPa) (Seed and De

Alba, 1986).

06

V 0'

Z 03

>

u

o

o.i

-11-

M> 7.3 orthquaku

% riots >33 213 slO 13

Ojolmm) 01 0.2 0.2302304 08

le/Ngo 3.3 41 4.4 4,4 4 33

_1_

X

4080120160200

Modifiad Con* P*n*tralion R*iilanc*. qe|-lf

240

Figure 18. Relationship between stress ratio causing liquefaction and cone tip resistance for sands and

silty sands (1 tsf100 kPa (Seed and D e Alba, 19 86).

There are several problems in applying the Seed

et al semi empirical approach:

(a) It has been developed for level or near

level ground conditions. For most dam

applications it can therefore only be

directly applied to assess whether

liquefaction would occur without the

dam.


47


55/127

Seed and Harder (1990) recommend

that to allow for the static driving stress

due to the dam, a correction factor

should be applied to the calculated

(Tav/CT'0). This is calculated using

(xayCT'0)a^=(xav/a'0)a^ Ka

where Ka is a correction factor

determined from Figure 19. To

determine Ka, the relative density and

a (the ratio of static driving shear stress

on a horizontal plane to the initial

effective overburden stress, (tav/a'0) has

to be determined, eg. from finite

element analyses.

The correction only applies to soils

where Tav/a'0


56/127

(c) In most natural soil deposits the SPT

values are variable (over that increase

which occurs with the increasing

overburden stresses). There are no

clear guidelines given by Seed et al as

to how this should be accounted for.

US SR (1989 (a)) discuss this issue and

present some useful practical points

which are recommended for use. These

include:

The results of each interval of each

drill hole, with regard to liquefaction

potential, should be prepared in a

table and should be presented on

geologic cross sections and profiles

to allow examination of the

frequency and continuity of those

intervals indicating liquefaction

susceptibility. From such a

presentation a judgement is drawn

as to whether or not the continuity

of potential liquefaction intervals

indicated is great enough to be of

concern.

If the deposit being sampled is

known to contain, or may contain

gravel, these coarse particles may

increase the blow count, implying a

more dense, less potentially

liquefiable soil. To check for this,

USBR (1989(a)) recommend

recording SPT blow counts for each

300mm of penetration, and

correcting for the erratic effects of

gravel (as a minimum, the three sets

of 150mm blow counts should be

checked to see whether this potential

irregularity is present). Where

gravel is extensive, shear wave

velocity methods should be used to

assess liquefaction potential.

X

}*

*x

V

X

X

X

5s .* *

X

t

X

35%

=0 ifFC


58/127

It is important to note that a positive result from

the Seed method only indicates that a soil may

be susceptible to liquefaction. It does NOT

mean that just because (part of) the dam

foundation will liquefy, the dam will fail. What

should follow such an outcome, is an

assessment of the extent and continuity of

potentially liquefiable soil, and the residual

undrained strength of that soil, for use in post

liquefaction analysis.

Liao, Venziano and Whitman (1998) provide

some useful information which will help in the

application of liquefaction in a probabilistic

framework.

5.3.4 Shear Wave Velocity Methods for

Assessing Liquefaction Potential

Shear wave velocity is affected by many of the

variables which influence liquefaction, eg.

(relative) density, confining pressures, stress

history, geologic age. Hence, it has some use as

an indicator of potential for liquefaction.

The shear wave velocity may be obtained by

downhole, crosshole or surface to downhole

seismic methods, or by a seismic cone

penetration test (a modification of the

piezocone test) developed by Campanella et al

(1986).

As discussed above, some coarse grained

cohesionless materials (gravels, cobbles)

suspected of being potentially liquefiable

cannot be successfully sampled using SPT. If

crosshole shear wave velocity data have been

obtained on the materials, and these data

accurately represent the deposits in plan and

section, then they provide a viable means for

making a judgement on liquefaction potential.

USBR (1989) recommend that:

if shear wave velocities are >365m/s, the

deposits may be judged non liquefiable

if shear wave velocities are between 245 and

365m/s, the deposit may be considered

likely to be non liquefiable, but supporting

evidence should be obtained

if shear wave velocities are < 245m/s, the

deposit may be judged liquefiable.

In practice, this will leave many sites in the

"grey" zone.

Shear wave velocity may also be used to assess

liquefaction by relating peak ground

acceleration, and a history of performance of

sites in earthquakes. Bierschwale and Stokoe

(1984) and USNRC (1985) give details.

US N RC (1985) also gives details of a method

which assesses peak strains due to the

earthquake from the shear wave velocity, and

compares this to threshold strain. Brief details

of these methods are given in Fell et al (1992).

5.3.5 Determination of Residual Undrained

Strength

In recent years there has been quite a lot of

discussion of the post liquefaction condition.

This is usually discussed in terms of "residual

(undrained) strength", "field residual strength",

or "steady state undrained strength". Some of

these are expressed as plots of residual

undrained strength versus SPT 'N' value (eg.

Seed (19 87), Lo and Klohn (1990), Seed and

Harder (1990). These plots are developed from

backanalyses of liquefaction failures. Figure 22

is the plot presented by Finn (1993).

Finn (1993) indicates that lower bound

strengths are often used for these analyses,

although 33rd percentile values have sometimes

been used. In either case, the lower bound or 33

percentile gives very low (-zero) residual

undrained strengths at SPT less than about N =

6.

An alternative approach adopted by a numb er of

authors including Ishihara (1994) and Finn

(1993), is to relate the normalised residual

undrained strength Su/ct'vo (where SU5 = residual

undrained strength and ct'Vo effective vertical

stress) to either SPT (N,)^ values (N values

corrected to 100 kPa effective stress, 60%

energy ratio hammer) or to cone penetration

resistance qc.


51


59/127

2000

I60C

H

O

Z

tu

cr

e

1.3 and maximum

compressive stress


87/127

Figure 34 Sequence of Analysis for Operating Basis Earthquake

(Guthrie 1986)

The preceding method could probably be used

for arch and buttress dams as well. However,

consideration would have to be given to the

appropriateness of the damping ratios and

corresponding limiting tensile strengths used.

(c) Non-linear Dynamic Analysis

A non-linear dynamic analysis of a concrete

dam is a complex analysis and would normally

be undertaken by specialist numerical analysts

experienced in such work. It would normally

be done only for major dams where the cost of

the new dam or the cost of remedial works for

80



88/127

an existing dam is sufficient to justify the

greater expense of this type of analysis.

Provided peak tensile stresses (static +

dynamic) do not cause cracking in the dam,

then a linear analysis will suffice. When

cracking occurs, stresses will re-distribute.

Therefore, when cracking is expected or there

are pre-existing cracks (e.g. open vertical

contraction joints) and the dam is essentially

3D m nature then a non-linear analysis should

be done.

There are three main ways in which cracks can

be modelled:

as distributed (smeared) cracks (zero elastic

modulus in direction perpendicular to

cracks)

as discrete cracks at finite element

interfaces

using fracture mechanics.

The first of the alternatives is computationally

the simplest. Further details on cracking are

given by Zienkiewicz, Valliappan and King

(1968), Rashid (1968), Mohraz, Schnobrich

and Gomez (1970), Darwin and Pecknold

(1978), Phillips and Zienkiewicz (1976),

Bazant and Cedolin (1979), Bazant and Ob

(1979), Argyris, Krempl and William (1977),

Gerstle (1981), Kotsovos and Newman (1978),

William and Warnke (1975), Cedolin, Crutzon

and Dei Poli (1977), Bicanic and Zienkiewicz

(1983), Zienkiewicz, Fejzo and Bicanic

(1983), Zienkiewicz, Hinton, Bicanic and

Fejze (1980), Pande and Shen (1982), Pal

(1974) and Chapuis, Rebora and Zimmermann

(1985).

A non-linear analysis will by its nature,

require a time-history analysis. Consideration

will have to be given to:

the way the compressibility of the storage

water is modelled

the way seepage pressures particularly

those due to water penetrating cracked

zones, are modelled.

The non-linear analysis of concrete dams is

still a developing and specialised field. Other

relevant papers include Waggoner, Plizzari

and Saouma (1993), Gao Lin, Jing Zhou and

Chuiyi Fan (1993), Greeves and Taylor

(1992), Cervera, Oliver and Galindo (1992),

Jing Zhou and Gao Lin (1992), Clough and

Ghanaat (1993), Fenves and Mojtahedi (1993).

7.3.4 Analysis of Permanent Deformations

In some concrete gravity dams subjected to

say the maximum credible earthquake, it may

be permissible for the dam to slide on its base

or within the foundations and be permanently

deformed after the earthquake. This of course

assumes that during or after the deforming

process, the security of the storage is not

preudiced allowing for the potentially

increased uplift pressures and lower strength

which may apply. Researchers such as Chopra

and Z hang (1991), Leger and Katsouli (1989)

and Danay and Adeghe (1993) discuss

calculations which indicate that typical

permanent displacements for large concrete

gravity dams subjected to earthquakes having

peak ground accelerations the order of 0.5g

can range from tens of centimetres to more

than half a metre. However, some dams with

suitable foundations would be able to

withstand small displacements. Dams relying

on post-tensioned ground anchors or drain

holes for normal load static stability, might not

be able to withstand these sort of movements

if the displacement was sufficient to shear the

anchors. However it may be acceptable to

have the anchors sheared for a low probability

earthquake provided the dam was stable under

the post earthquake load case.

The type of analysis required to compute

permanent deformations is similar to that

carried out for embankment dams i.e. Makdisi

and Seed (1978). The dam is considered as a

block with a limiting sliding strength along its

foundations. The dam is subjected to a time

varying input of acceleration. When the

acceleration is greater than the limiting

acceleration (the acceleration causing inertia

forces which are greater than the sliding

strength of the foundations) the dam will move

on its foundations. The parts of the

accelerogram greater than the limiting

acceleration are double integrated to obtain

cumulative displacements.

The type of analysis just described is a

requirement of the US Corps of Engineers

method for dynamic analysis of concrete

gravity dams when the sliding safety factor is

less than one. The sliding analysis is carried

out for horizontal planes through the dam

where there is cracking. A crack is assumed

through the dam with suitable slip elements

along the crack interface.


81


89/127

In their study of base sliding, Chavez and

Fenves (1993) found that sliding was not

likely to occur if the storage is less than half

full. Other conclusions were :

vertical ground motion has almost no effect

on the sliding displacement (it slightly

increased the maximum stresses). This is

partly because the vertical and horizontal

ground motion do not necessarily coincide.

the assumption of rigid foundation rock can

significantly overestimate the amount of

base sliding

At the present stage of development of the

computation of permanent deformations and

the determination of suitable maximum

displacements there is still much work to be

done especially regarding local conditions.

These guidelines therefore counsel that

considerable care be used if a dam is to be

allowed a permanent deformation following an

earthquake. Adequate sliding and overturning

stability must exist after the earthquake using

foundation strengths appropriate to the

displacement (usually the residual strength)

and uplift appropriate to the displaced

condition, allowing for opening of joints and

bedding, and for reduced (or no) drainage

capacity.

7.4 Design Earthquake and

Hydrodynamic Loads

7.4.1 Earthquake Parameters

As discussed elsewhere in these guidelines,

dynamic analysis can be carried out in the

frequency domain or the time domain. For the

former, response spectra are required while for

the latter, accelerograms are required.

(a) Response Spectra

A response spectrum shows the extent to

which any single degree of freedom structure

with an assumed level of damping would

respond to particular earthquakes.

Knowing the natural frequencies of vibration

and the corresponding mode shape for a

structure, the spectral accelerations

corresponding to particular natural frequencies

and damping ratios can be converted to inertial

loads.

The response spectrum used in the analysis of

a dam should be site specific and relate to the

peak ground accelerations examined. The

response spectrum should also reflect the

frequency mix and duration of the design

earthquakes. It will therefore be derived from

a number of earthquakes having various

epicentral distances from the site and

consequently, different acceleration

attenuation functions. The response spectra

should be obtained from a seismologist as part

of the assessment of seismicity of the dam site.

(b) Accelerograms

Where a time-history analysis is to be done, at

least three different accelerograms appropriate

to the dam site and for a particular peak

ground acceleration, should be used. These

accelerograms may be recorded accelerograms

which are suitably scaled (accounting for

change in frequency mix and phase with

change in peak ground acceleration) or

synthetic accelerograms which fit the response

spectra for the site. Care should be taken in

selecting accelerograms which are similar to

Australian earthquake conditions, and advice

should be obtained from a seismologist.

7.4.2 Hydrodynamic Pressures

Any movement of the dam and foundation will

cause movement in the water of the storage

and in turn, the pressures generated by the

water will impose forces on the dam.

Engineers have traditionally used

hydrodynamic pressures derived by

Westergaard (1933). These pressures are

commonly converted into equivalent lumped

'virtual' masses which are attached to the dam.

Westergaard's pressure distribution assumes

that the water in the storage is incompressible

and that the dam and its foundations are rigid.

However, this is not always so. In high

gravity dams and slender arch dams especially,

where the dam is flexible, there can be

considerable interaction or coupling between

the dam and the storage.

Considerable work on the interaction of

gravity dams and their storages has been done

by Chopra and his fellow researchers. Details

of the work are given in Chopra (1967),

Chakrabarti and Chopra (1974), Chopra,

Chakrabarti and Gupta (1980), Chopra and

Gupta (1981). Other relevant papers include

82

AN CO LD Guidelines for Design ofDams for Earthquake


90/127

Clough and Chang (1980), Dungar (1978),

Hall (1988), Zienkiewicz, Paul and Hinton

(1983) and Tsai and Lee (1989).

Besides the lumped, 'virtual' mass approach,

the interaction of a dam with the water in its

storage can be determined by treating the

water as a 'solid' having zero shear modulus

but retaining compressibility. This approach,

although simple in principle, has a number of

numerical problems. An alternative approach

is therefore preferable where from the

beginning, the shear components of stress in

the fluid are neglected. This latter approach

includes the effect of water compressibility. It

also will allow the limiting case of

incompressible water to be considered.

Accounting for water compressibility and

coupling of a dam with the water in its storage

adds considerable computational effort to

determining the effects of earthquake loading

on a dam. Neglecting coupling and water

compressibility in the simpler "added mass"

approach is not considered significant for

excitations at frequencies below the natural

frequency of the reservoir.

An estimate for the fundamental frequency of

a reservoir can be obtained from:

f =__

w4H

eff

where: fw = the fundamental frequency

C = the compression wave speed

in water (l,439m/s)

Heff = an effective depth of the

reservoir.

The above relationship was obtained from

Duron, Ostrom and Aagaard (1994) and

applies strictly to an infinite reservoir of

constant cross section. Duron and Hall (1988)

indicate that if the ratio of to the

fundamental frequency of the dam-foundation

alone (fj) is near unity, water compressibility

will have a significant effect. For ratios of fw

to fj much greater than one (e.g. >1.5)

incompressible fluid behaviour can be

assumed.

7.4. i Uplift/Seepage Pressures

The USBR (1977) considers that the uplift

pressure within the crack is zero while ICOLD

(1986) assumes full headwater pressure but

recognises the need for further research.

Guthrie (1986) uses the pre-crack uplift

pressure diagram.

These guidelines recommend that for the

duration of the earthquake, the pre-earthquake

uplift pressure distribution is used for the

stability analysis. However for the post

earthquake analysis, consideration should be

given to the amount of cracking and the post-

earthquake efficiency of the dam's drainage

system. In the post-earthquake situation, full

headwater pressure is assumed to exist in a

crack at least as far as the line of drains. If the

drains have sufficient capacity and they have

not been disrupted by sliding of the dam, then,

if the crack extends, past the drains, a

significant reduction in uplift pressure should

be considered. Typically, the pressure at the

line of drains in this case might be the

tailwater pressure plus 50% of the difference

between headwater and tailwater pressures.

The pre-earthquake uplift pressure distribution

might have had a 67% reduction. The lesser

reduction for the post-earthquake case allows

for the greater amount of drainage with which

the drains would have to cope.

7.5 Design Criteria

7.5.1 General Approach

The working group has had two major

difficulties in preparing suitable design criteria

for concrete gravity dams.

(1) The current ANCOLD (1991)

guidelines for design criteria for concrete

gravity dams are based on a limit state

approach with partial factors of safety. This

method has proven to be difficult to use on

dams subject to significant earthquake loads.

The ANCOLD (1991) guideline is under

review to address these problems. In the

interim, it is recommended that the design

loadings and acceptance criteria described

herein are used. These are based largely on

the BC Hydro (1995) guidelines.

(2) Existing guidelines for design of

concrete gravity dams are not simply applied

to a risk based approach. As a result it has not

been practicable to develop these guidelines to

directly apply to a risk based approach, and

they are given in terms of a deterministic

method using OBE and MDE. For

completeness, flood and static load case are

also listed. Those wishing to use a risk based


83


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approach may do so, taking account of the

general intention of the load cases detailed

herein.

7.5.2 Loads and Load Cases

The loads considered in the assessment of

concrete gravity dams and their foundations

should include the following:

Dead loads of permanent structures,

equipment and foundation rock (D).

Water load due to maximum normal

headwater level combined with the most

critical concurrent tailwater level (H).

Water load due to maximum flood

headwater level based on the Inflow Design

Flood (IDF) with corresponding tailwater

levels (H).

Foundation uplift (U), both at the concrete-

rock contact and at critical discontinuities

within the foundation.

Static and dynamic thrust created by an ice

sheet, for reservoirs subject to freezing (I).

Vertical and horizontal loading due to rock

or soil backfill (both natural or engineered),

including potential effects of liquefaction

and loads from silt deposited against the

dam (S).

Load due to Operating Basis Earthquake

(OBE) Q'

Load due to Maximum Design Earthquake

(MDE ) (Q).

Determination of the loads should take into

account the actual field conditions and

instrumentation records.

Foundation uplift assumptions should reflect

the stress state and condition of the dam and

foundation. Disruption of the dam and/or

foundation condition due to an earthquake

should be recognised in assessing the uplift

assumptions for the post-earthquake case.

The dam and foundation should be assessed

for the following load cases:

(a) Usual Load Case

permanent and operating loads should be

considered for both summer and winter

conditions including self-weight, ice (where

applicable), silt, earth pressure, and the

maximum normal reservoir level with

appropriate uplift pressures and tailwater level.

(D + H +1 + S + U)

(b) Unusual (Flood) Load Case

Permanent and operating loads of the Usual

Load Case, except for ice loading, should be

considered in conjunction with reservoir and

tailwater levels and uplift resulting from the

passage of the IDF.

(D + H' + S + UF)

where subscript "F" refers to the flood case.

The potential should also be assessed for

reservoir levels higher than would result from

passage of the IDF, such as those due to

operating failures or other unusual conditions.

The effects of ice loads should not be

considered simultaneously with flood

conditions.

(c) Unusual (Earthquake) Load Case

Permanent and operating loads of the usual

load case except for ice loading, should be

considered in conjunction with earthquake

loading associated with the Operating Basis

Earthquake (OBE). The effect of ice loads

should not be considered simultaneously with

OBE earthquake conditions.

The analysis should be carried out for the dam

empty case.

(d) Extreme Load Case

Permanent and operating loads of the Usual

Load Case should be considered in

conunction with seismic loads of the

Maximum Design Earthquake (MDE).

(D + H + S + Q + U)

The effects of ice should be given special

consideration, recognising the high uncertainty

associated with ice loading on earthquake

loading, and its effects on the dam.

(e) Other Load Cases

Where earthquake-induced cracking at the

concrete-rock interface or any weak section is

identified, a stability analysis should be

carried out to assess whether the dam in its

post-earthquake condition is capable of

resisting loads of the Usual Load Case.

84



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(D + H + S + UpQ)

where subscript "pg" refers to the post-

earthquake case.

Concurrent ice loading (with the post-

earthquake condition) may be considered in

areas where appropriate.

A landslide generated wave case should be

considered where existing or potential

landslides, which may affect the reservoir,

have been identified. Combinations with other

loads would be site specific.

An inoperative drain case assuming plugged

drains may be assessed and could be

considered as an Unusual Load Case.

7.5.3 Acceptance Criteria

All kinematically feasible failure modes,

analysed by the single-slice, rigid-body, force

equilibrium method, should satisfy the

acceptance criteria shown in Table 18.

Table 18

Stability Index Acceptance Criteria

Load

Sliding Factor

(Note 1)

Position of Resultant Force

(Note 2)

Minimum Compressive

Stress Factor

(Note 3)

Usual

1.5-2.0

Mid-third of surface

No tension

4.0

Unusual (Flood)

and

Unusual (OBE )

1.3-1.5

Mid-half of surface

One-quarter tension

2.7

Extreme (MDE)

1.1 - 1.3

Within surface

1.3

Post-earthquake

1.2-1.4

Mid-half of surface

One-quarter tension

2.7

Notes: 1. Sliding Factor (Frictional Analysis) = Resisting Forces

Applied Force

Lower values of the range apply where the geology and the strength parameters are

reasonably well known.

2. Vector summation of all forces, including uplift, acting on the analysis surface

3. Compressive Stress Factor = Unconfined Compressive Strength

Compressive Stress Normal to Surface

To be considered primarily for massive but low strength rock and weak deteriorated

concrete.

7.5.4 Post Earthquake Stability

If a dam is likely to be severely damaged after

being subjected to the MDE, considerable time

may elapse before the dam can be repaired or

the storage lowered. Consequently, all parts of

the dam will need to remain stable after an

extreme earthquake event. The stability of the

dam should therefore be checked for static

loading conditions. The assumed uplift

pressure distribution should be as discussed in

sub-section 7.4.3, i.e. full headwater pressure

within cracks emanating from the upstream

face.

7.5.5 Foundation Stability

Where there is the possibility of a sliding

failure along faults, shears and/or joints, the

stability of the foundations should be


85


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examined. This examination should be made

for both the earthquake and post-earthquake

cases. The dam itself should be examined for

local overstressing due to foundation

deficiencies.

Sliding failure is especially likely when

discontinuities and/or horizontal or sub-

horizontal seams close to the foundation

surface contain clay, or have been previously

sheared eg. bedding surface shears due to

stress relief, folding, or associated with faults.

7.6 Dynamic Material Properties

7.6.1 Concrete

The compressive and tensile strengths of

concrete increase with increased rate of

loading. The dynamic compressive and tensile

strengths of concrete can therefore be expected

to be greater than the static strengths. As

dynamic compressive stresses are rarely of

concern, the allowable compressive stress for

static loading can be used also for dynamic

loading.

Raphael (1984) states that the apparent tensile

strength of concrete under seismic loading

which should be used with linear finite

element analyses is given by:

fr =065 fc 2P

where fc is the concrete compressive strength

in MPa and fr is the apparent seismic tensile

strength in MPa. Values given by this formula

are some 50% greater than the apparent tensile

strength for static loading. Raphael suggests

that fr is twice the splitting strength of the

concrete under static loading.

C lough and Ghanaat (19 93) suggest that the

apparent dynamic tensile strength is about

25% greater than the measured static value

which gives apparent tensile strength about

20% of the standard compressive strength.

They further suggest that there may be a 15 to

20% loss of strength across lift joints. These

figures may be even lower for poorly

constructed or defective lift surfaces,

fiowever, the peak dynamic tensile stresses

only exist during a fraction of a response

cycle. Even though these peak stresses may

greatly exceed the tensile strength of the

concrete, any cracking that might be initiated

will not have time to fully develop. It is well

recognised that a single spike of excessive

localised tension should not be taken to

represent dam failure.

In consideration of the above however, these

guidelines recommend that for sound lift

surfaces, the apparent tensile strength to be

used is 16% of the standard compressive

strength.

For dynamic modulus of elasticity, Clough and

Ghanaat (1993) suggest a value 25% greater

than the static value and these guidelines

recommend this be adopted. For existing

dams, the elastic modulus of the concrete mass

may be determined using geophysical means

(e.g. derived from measured shear wave

velocity). Values obtained should be

compared with static and dynamic small

sample laboratory test values for credibility.

7.6.2 Rock

In most cases the stability of the dam will be

controlled by sliding in or on the dam rock

foundation. To carry out static and dynamic

analyses, it will be necessary to:

assess and map surface exposure,

excavations and drill core to define the

geology of the site. In particular the 3-

dimensional orientation, continuity and

detailed nature of bedding, joints, and other

features such as shears are required.

the shear strength of the foundation rock

should be determined using appropriate

rock mechanics techniques, such as those

described in Hoek (1983, 1990, 1994),

Hoek and Brown (1980), Patton (1966),

Barton and Choubey (1977), Barton and

Bandis (1991). These require an

assessment of the orientation, spacing,

continuity, shape and roughness of

discontinuities in the rock (e.g. joints), and

the strength of the rock substance as this

varies with confining stress. Care should

be taken in applying these techniques, to

account for the presence of continuous,

adversely oriented low strength surfaces

such as bedding surface shears, faults or

shears, and to take account of the

mechanisms of failure.

The Hoek, and Hoek and Brown methods give

strengths for "undisturbed rock" and

"disturbed rock". The latter should generally

be adopted unless advice from an experienced

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rock mechanics specialist indicates otherwise,

because of the uncertainty as to the accuracy

of the Hoek methods, and since the

undisturbed rock strengths apply to confined

conditions such as underground openings,

where dilation or shearing causes increases in

normal stress.

It will be noted these stress methods give non

linear failure envelopes, with high friction

angle and low cohesion at the normal stresses

usually applicable to dams. They also allow

estimation of the modulus of the rock mass

(Hoek, 1994). The dynamic modulus may be

higher than the static modulus as discussed in

Clough and Ghanaat (1993) and Scott and Von

Thun (1993) and may, as for concrete, be

obtained by geophysical means, or by relation

to the static modulus.

A dam's foundations will normally contain

joints, shears, and bedding. Consequently, it

will not be possible to transmit tensile stress

within the foundations and the allowable

tensile strength for the foundations will

therefore normally be assumed to be zero.

However, if extensive site investigation and

strength testing is able to prove that the

foundations for a particular dam site are

capable of transferring tensile stresses, then

the tensile strength of the rock may be

included.

Where foundation rock strength becomes

critical, as they often will, advice should be

obtained from a person expert in rock

mechanics.

8. APPURTENANT

STRUCTURES

8.1 Introduction

A number of subsidiary structures associated

with a dam are essential for the dams

operation. Consequently damage to or

destruction of these appurtenant structures

would be prejudicial to the dam's safety. An

important facility at a dam is one that allows

water to be released in a controlled manner. If

therie has been an earthquake and the dam is

damaged to the extent that the dam is not

serviceable then it may be necessary to lower

the storage so that remedial works can be

undertaken. It will therefore be necessary that

not only the outlet structures and their gates

and valves remain serviceable but also that

there is proper access to these structures.

Bridges and roads may need to remain in a

sound state after an earthquake depending on

their importance.

Generally, appurtenant structures should be

such that:

they maintain their normal operating

condition after an operating basis

earthquake

they are not damaged to an extent where

they could allow sudden or uncontrolled

loss of water from the storage for a more

extreme earthquake up to the maximum

design earthquake.

8.2 Performance Requirements

This sub-section gives the performance criteria

for the operating basis earthquake (OBE) and

the maximum design earthquake (MDE) which

could be the maximum credible earthquake

(MCE). Performance requirements are given

for a number of appurtenant structures

including intake towers, outlet conduits, outlet

works, spillway gates, spillway piers, spillway

gate hoist piers, access bridges and piers,

access roads.

Intake Tower

OBE: Static and dynamic loads to

induce maximum concrete and

steel reinforcement stresses

which satisfy AS3600

(Concrete Structures Code)

(i.e. limited amount of

reinforcement yielding) and

the tower and its base remain

stable.

MDE: Significant amount of

reinforcement can yield

horizontal reinforcement

designed to prevent vertical

reinforcement from buckling

and to contain concrete when

it is in compression (i.e.

concrete contained between

inner and outer layers of

vertical reinforcement will not

spall away).

Outlet Conduit


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O B E :

MDE:

Outlet Works

O B E :

Static and dynamic loads to

induce maximum concrete and

steel reinforcement stresses

which satisfy AS3600

(Concrete Structures Code).

Dynamic loads to include

those induced from the

earthquake effect on the

overlying dam.

Conduit not to collapse or

rupture - Collapse could lead

to undermining and

subsequent failure of an

overlying embankment.

Rupture could cause piping or

destabilise an embankment by

a marked increase in pore

pressure.

All valves to maintain their

normal operating capabilities.

MDE: Emergency closure and

regulating valves (especially

low level release valves) to

maintain operating capability -

storage may have to be

quickly lowered if parts of the

dam are damaged and need

remedial works or relief of

hydrostatic loads.

Spillway Gates

OBE: Gates to maintain normal

operating capability.

MDE: Gates retaining permanent

storage at the time of an MDE

should not fail to the extent

where water from the storage

is released in an uncontrolled

manner. MDE should not

cause the gates to distort to an

extent that they cannot be

opened or closed.

S pillway Piers

O B E :

Carry out appropriate dynamic

analysis for the spillway piers

for earthquake loading in the

upstream/downstream and

transverse directions.

Combined static and dynamic

loads should satisfy ASS 600


Earthquake loads from the

orthogonal directions may be

combined on a square root of

the sum of the squares basis.

MDE: Carry out dynamic analysis as

for the OBE. Combined static

and dynamic loads may cause

cracking but piers must

remain stable for overturning

and sliding. Piers should not

be permanently displaced to

the extent where the spillway

gates become jammed.

Spillway Gate Hoist Piers

OBE: Carry out appropriate dynamic

analysis similar to that for

spillway piers. Combined

static and dynamic loads

should satisfy AS3600


MDE: If the gates can be operated

from an alternative position

(possibly, in a less efficient

manner), then the spillway

gate hoist piers (and hoist

bridge) can be allowed to fail.

If the continuing operation of

the gates depends on the

continued viability of the hoist

bridge piers (and hoist bridge)

then carry out an appropriate

dynamic analysis similar to

that for the spillway piers.


loads may cause cracking but

the piers must remain stable

for overturning and sliding.

Note: theappropriate

dynamic analysis

required for spillway

piers and spillway

hoist piers may be part

of an overall dynamic

analysis for a concrete

dam.

Access Bridge arid Piers

OBE: Carry out appropriate dynamic

/ of the pier and bridge

'The anaysis may be

88

AN CO LD Guidelines for Design

Wke'

*4


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part of an overall dynamic

analysis for an intake tower.

The pier and bridge system

should be examined for an

earthquake in the direction of

the bridge access and

perpendicular to the bridge

access. Earthquake loads may

be combined on a square root of

the sum of the squares basis.


loads should satisfy the

appropriate Structures Code).

MDE: If there are alternative means

of access then the access bridge

can be allowed to fail. If there

are no alternative means of

access then the bridge and piers

should remain capable of

carrying their design loads.

Access Roads

OBE: Access roads to the dam and

its appurtenant works should

remain passable immediately

after the OBE. Therefore any

likely land slip areas along the

access roads should be checked

for stability. There should be no

land slips either during or after

an OBE.

MDE: Access roads to the dam and

its appurtenant works may

become impassable during or

immediately after an MDE.

However, they should be easily

cleared. Therefore, while there

may be land slips onto the

roads, the roads themselves,

should not be allowed to

collapse where there is no easy,

alternative access route.

8.3 Intake Towers

8.3.1 Analysis

The analysis method for intake towers is

described here in general terms only. A more

complete description of the method on which

the following general description is based, is

given by Chopra and Goyal (199 1) and Goyal

and Chopra (1989).

The method presented here is based on a

simplified dynamic analysis in the frequency

domain using a suitable response spectrum. It

uses an added mass representation of

hydrodynamic effects due to surrounding

(outside) water and contained (inside) water

(in the case of wet towers). In addition, it

includes the effects of tower/foundation

interaction.

The steps of the method are:

(i) Select suitable response spectrum.

(ii) Compute the added hydrodynamic

mass of water using Goyal and Chopra

(1989).

(in) Determine the structural properties of

the tower

mass per unit height

flexural stiffness

modal damping ratios.

(iv) Compute natural periods and mode

shapes for the first two modes of

vibration.

(v) Determine the spectral accelerations

for the first two modes of vibration

from the response spectrum.

(vi) Compute the generalised mass and

generalised excitation terms using the

mass distribution and the mode

shapes.

(vii) Compute the inertia forces using the

spectral accelerations, generalised

mass and excitation terms, mass

distribution and mode shapes.

(viii) Add the inertia loads from the first

two vibration modes on a square root

of the sum of the squares basis.

(ix) Compute bending moments and shear

forces from (viii).

(x) Design tower according to AS3600


8.3.2 Design Criteria

As discussed in Sub-section 8.2, an intake

tower is designed elastically for the OBE.

Generally, it will not be necessary for the


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tower to behave elastically during the MDE.

In order to ensure that the tower can survive

intense ground shaking due to the MDE with

limited damage, it should possess a ductility

capacity greater than the ductility

requirements imposed by the ground motion.

A suitable method for this is described in

Chopra and Liaw (1975) where a ductility

factor of two is recommended (ratio of

maximum permissible displacement to the

yield displacement).

/

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R E F E R E N C E S

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guidelines for design of dams for earthquake

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