pc frame building components for westin hotel part 1

Spr ing 2009 | PCI Journal84

Editor’s quick points

n This research and development effort examines the design and performance of elements required for the successful con-struction of a precast concrete building system that emulates flat-slab construction.

n The research confirmed the flexural and connection capacity of the rib slab, the flexural and shear capacity of the precast concrete column capital, and the moment-resisting capacity of the precast concrete column connection.

n A lack of stiffness with the column-to-column connection was found, and solutions were developed for this connection.

Precast concrete building system components for the Westin Resort Hotel, part 1: Experimental validationJohn Hanlon, Charles W. Dolan, David Figurski, Jiangang Deng, and J. Gregory Dolan

Rocky Mountain Prestress Corp. proposed an alternative design for the construction of a nine-story, flat-slab build-ing in Avon, Colo., using a total–precast concrete building structural system. The building system varies significantly from the systems evaluated in the Precast Seismic Struc-tural Systems (PRESSS) program and required supplemen-tary experimental evaluation to demonstrate performance.1 This system consisted of five basic elements: a precast concrete stair/elevator core; a 10-in.-deep × 4-ft-wide (250 mm × 1.2 m), prestressed concrete beam-slab unit; a 10-in.-deep, prestressed concrete, rib-slab floor element; a 10-in.-thick, variable-width beam slab; and an integrated precast concrete column and column capital.

The development of an integrated column capital allowed each column to be erected independently and the beam slabs and rib slabs to be installed on the column capital. This design allows the precast concrete building system to emulate post-tensioned, flat-slab construction. The ability to provide a precast concrete column capital of varying dimensions allows the design to be directly substituted for a flat-slab project and for designers to incorporate balco-

85PCI Journal | Spr ing 2009

Figure 2. A load test of a rib-slab panel with dapped ends was conducted at the Rocky Mountain Prestress plant in Denver, Colo.

Figure 1. This schematic plan shows a precast concrete building using the components described in this paper.

Legend

Column and capital Beam slab

Rib slab

Building perimeter


nies, recesses, and variable perimeters (Fig. 1). “Precast Concrete Building System Components for the Westin Resort Hotel, Part 2: Design and Construction” will discuss the implementation of the system.

Initial research and development examined the beam-slab and rib-slab units. The beam slab was a 10-in.-deep × 4-ft-wide (250 mm × 1.2 m), solid prestressed concrete slab that connected to the column capital with a Cazaly hanger.2 The hanger provided both the vertical reaction and, when field welded, a construction diaphragm connection.

The rib slab is a modified prestressed concrete double-tee shape with a 2-in.-thick (50 mm) concrete slab and 8-in.-deep × 15-in.-wide (200 mm × 380 mm) ribs, for a total depth of 10 in. Tests to 100% of the ultimate design load capacity on the rib slab with a dapped end connection established for the Avon project confirmed the feasibility of the basic floor system (Fig. 2). The hanger replaced the dapped end because it uses the full section depth, eliminates the need for bearing ledges, provides the diaphragm tie, and simplifies production. The hanger connection was proof tested to over 100% of the ultimate design load. A cast-in-place concrete topping covered the hangers and the welded-wire reinforcement, which provides the final diaphragm.

The critical element in the building system is the integrated column-capital unit. This research project provided proof of the concept for this structural system and identified con-struction or structural issues needing further development.

The test program demonstrated the fundamental behavior of the precast concrete structural system and its compo-nents, verified erection methodologies, and confirmed that the column-capital connection was adequate for the proj-ect. Stiffness deficiencies in the initial match-cast column connection were corrected prior to production.

Column connection

The column-to-column connection required a full- or partial-moment-resisting detail that could be quickly installed in the field. Full- or partial-moment-carrying capacity allowed the column elements to be placed rapidly with minimum time on the crane hook and to remain stable for the installation of the beam slabs with little or no brac-ing. The initial design placed four Grade 75 (517 MPa) no. 10 (32M) Dywidag threaded bars in each corner of the column. The bars were aligned in the form, connected with a threaded coupler, and match cast to the adjacent piece. After casting, the coupler was backed off and the elements were separated. This procedure ensured that the threaded ends aligned in the field and that the field connection was completed by rethreading the couplers (Fig. 3).

A lock nut on either side of the coupler secured the connec-tion. A steel bearing plate assisted alignment of the column

ends. The three elements stacked together demonstrated that a multipiece system could be assembled as designed. Concerns remained about the stiffness of the connection and the feasibility of field assembly of the connection with an eccentric slab load.

Full-scale test specimen

To evaluate the full-scale installation and strength of the column-capital element, an L-shaped test frame was fabricated and a full-sized column-capital element and a match-cast column stub extension were cast (Fig. 4). The test frame had a reaction wall on the west side of the col-umn to apply horizontal loads. The column test specimen was cast with a 27 in. (685 mm) capital extension beyond the face of the column on the west side and a 51 in. (1.3 m) extension on the east side. The column was cast with a 2 ft (0.6 m) stub extension to replicate field installation detail (Fig. 5). The east capital loading points were offset from the theoretical location by ±1 in. (±25 mm) in each direc-tion to generate a small amount of torsion in the capital. This torsion replicated possible eccentric load effects from beam-slab connections.

The column and stub were match cast with full-length,

Figure 3. The photo shows the initial column-to-column connection detail.

Coupler

Lock nut


Field erection of eccentrically loaded specimen

The offset slab construction allowed assessment of erec-tion issues associated with an eccentric element. The first field test was to lift the piece without damaging the ex-tended splice bars, place the unit, and tighten the connec-

assembled threaded bars and couplers installed at the splice location. After the combined column was removed from the form, the bar couplers were backed away from the splice location. The stub and the column were then separated, and the stub was installed in the test frame and grouted in place.

Figure 4. This schematic illustrates the test setup for the full-scale test program. Note: PVC = polyvinyl chloride. 1 in. = 25.4 mm; 1 ft = 0.305 m; 1 ksi = 6.895 MPa.

6 ft

4ft

2 ft 6 in. 2 ft 6 in.

1 ft

1 ft

1 ft

1 ft

Four 3-in.-diameter PVC pipe sleeve

6 ft

12 in

.

Hydraulic cylinders

2 ft

Test frame base

8 in

.

Testframewall

6 in.

3 ft

Eccentric length

3-in.-diameter pipe

1-in.-diameter, 145 ksi thread bar

Coupling system


tors. A hand-powered come-along was attached to one of the lifting loops and tightened to the crane hook. Tighten-ing the come-along allowed the eccentricity to be adjusted so that the column would hang vertically (Fig. 6).

The couplers on the compression side of the connection engaged almost immediately. However, the eccentric-ity created a small separation on the tension side of the connection that was larger than the couplers could accom-modate. The piece was lifted off the connection, and the vertical alignment was corrected with the come-along. With the vertical alignment corrected, the couplers were tightened (Fig. 7). Once the procedure was established, the time for setting and coupler connection of the piece was about 10 minutes.

When the column capital was released from the crane, the piece tilted toward the eccentrically loaded east side due to the tolerance of 1/16 in. to 1/8 in. (1.5 mm to 3.0 mm) in the coupler threads. The eccentric load compressed one side and stressed the other side in tension, creating a connection with an opening on the tension (west) side of the match-cast joint. Tightening the coupler and the lock nuts did not reduce the gap. As the lock nuts were tightened, the bars were pulled against the threads in the coupler and the bars extended from the connector. This had the effect of lifting the columns apart and increasing the tilt. The design an-ticipated this behavior, but the magnitude of the deflection was larger than predicted.

Figure 5. Threaded bars apply loads to the full-scale test specimen.

Figure 6. The come-along positions the eccentric column-capital assembly as it is being lifted onto the test frame.


Figure 7. Workers complete the initial column connection.

Table 1. Predicted column-capital strength

Condition ACI equationPredicted nominal shear capacity

Vn , kip

Total punching shear (ACI 318-05, section 11-12) Vn = 4 fc

' 4 bc +d2

⎛⎝⎜

⎞⎠⎟d

⎡⎣⎢

⎤⎦⎥

287

Shear-flexure 2

fc' (ACI 318-05, Eq. [11-3])

Vn = 2 fc

' bd 63

Shear-flexure 2

fc' plus stirrups (ACI 318-05, Eq. [11-3]

and Eq. [11-15]) Vn = 2 fc

' bd + Avfyds

100

Shear-flexure 3.5

fc' plus stirrups (ACI 318-05, Eq.

[11-5] and Eq. [11-15]) Vn = 3.5 fc

' bd + Avfyds

148

Flexure n.a. 119

Failure load n.a. 115

Predicted failure load/predicted flexure load n.a. 0.97

Notes: Av = area of shear reinforcement = 0.40 in.2; b = width of compression face of member = 48 in.; bc = cross-sectional dimension of column core measured center to center of outer legs of the transverse reinforcement = 18 in.; d = distance from extreme compression fiber to centroid of longitudi-nal tension reinforcement = 9.3 in.;

fc

'

= specified compressive strength of concrete = 5000 psi; fy = specified yield strength of reinforcement; n.a. =

not applicable; s = center-to-center spacing of shear reinforcement = 8 in.; Vn = nominal shear strength. 1 in. = 25.4 mm; 1 kip = 4.45kN; 1 ksi = 6.895 MPa.


A threaded bar and horizontal jack were installed at the top of the column to pull the piece into vertical alignment (Fig. 4). Although this reduced the gap at the match-cast interface, it did not allow the connectors to significantly tighten the bars.

Overall, the authors concluded from the erection trial that the match-cast units could be erected efficiently and that the threaded connectors would align and could be tight-ened. The column was stable and safe. However, a method of tightening the connection was needed for effective mo-ment transfer, especially for multistory erection.

Column-capital strength and performance

The largest ultimate load applied to the column-capital unit was 80 kip (360 kN) for the Avon project. A shear load of 100 kip (450 kN) at the loading points was tested to validate the design methodology and details. This allowed validation for future projects with longer spans or higher loads.

The objectives of the column-capital test were to experi-mentally confirm the strength of the capital, verify that flexure—not shear behavior—limits performance, confirm which shear equation from the American Concrete Insti-

tute (ACI) Building Code Requirements for Structural Concrete (ACI 318-05) and Commentary (ACI 318R-05)3 would be applicable for the design, and explore the effects of unbalanced moment on the column-capital element. Five strength predictions were made for the specimen:

punching shear•

shear and flexure using a concrete capacity of 2• fc' ,

where fc' is the compressive strength of concrete

shear and flexure using 2• fc' plus no. 4 (13M) stir-

rups at 8 in. (200 mm) on center

shear and flexure using a variable • Vc for nominal shear strength provided by the concrete according to the ACI 318-05 Eq. (11-5) plus no. 4 (13M) stirrups

flexural capacity•

Table 1 lists the equations, nominal-strength predictions, and results. Assuming that the higher shear capacity of option 4 was used, the predicted nominal capacity of the column capital was 118 kip (525 kN), and the failure mode was predicted to be flexure.

A beam slab is not always placed symmetrical on the

Figure 8. The basket truck is used to observe the cracking on the column capital.


column capital. Therefore, one objective of the test was to apply an eccentric load to the column-capital assembly. An initial 25 kip (110 kN) load was applied on the west side of the specimen, simulating the eccentrically loaded slabs. The load was locked off, and the hydraulic cylinders were attached to pressure gauges. The west bars became pas-sive restraints as the full moment load was applied to the east side of the specimen (Fig. 8). The passive, preloaded hydraulic cylinders also served as load cells during testing. Cylinders and pressure gauges were calibrated before and after the testing.

Dial gauges were mounted on the threaded bars to measure deflection. The deflection readings were corrected for axial lengthening of the bars during loading. An unloaded, longi-tudinal threaded bar was placed through the column top and the test frame for safety in the event that the column suddenly failed. It remained unloaded for the duration of the test.

First cracking of the capital was noticed at a load of 28.9 kip (129 kN) versus the predicted cracking load of 29.2 kip

(130 kN). The initial cracks were less than 0.002 in. (0.05 mm) wide. The crack width remained less than 0.010 in. (0.25 mm) throughout the service-level-load range (Fig. 9).

First yielding of the reinforcement was noticed at about 95 kip (420 kN). The crack pattern indicated the initial formation of secondary compression failure at about 100 kip (450 kN) (Fig. 10). At this point the dial gauges were removed. The specimen failed at about 115 kip (509 kN) or 2657 kip-in. (298 kN-m) of moment, which was within 3% of the predicted flexural strength. There was no indication of shear failure in the specimen.

At failure of the east capital, the west-capital load was carrying 75 kip (330 kN), corresponding to a moment of 1548 kip-in. (173 kN-m). An unbalanced moment of 1109 kip-in. (124 kN-m) was carried by the column, which was about 40% of the theoretical moment capacity of the column. Thus, even with the large coupler tolerance, the joint was functionally capable of carrying the applied moment.

Figure 10. The photo illustrates the crack pattern on the column capital near the nominal capacity.

Side of capital Top of capital

Figure 9. The load-deflection curve for east capital shows the stiffness of the slab connection. Note: 1 in. = 25.4 mm; 1 kip = 4.45 kN.

0 5101520253035404550

0500

1000150020002500300035004000

0.00 0.10 0.20 0.30 0.40 0.50 0.60

Mom

ent,

kip-

in.

Deflection, in.

Late

ral l

oad,

kip


critical design condition, and no indication of punching shear was detected. The eccentric torsional effects were insignificant.

Column test program and observations

The original intent of the column connection test was to subject the column to a series of reversible loads up to 3% drift, consistent with the recommendations of ACI’s report ITG-1.14 for seismic applications. The tolerance of the column-to-column connection suggested that taking full hysteresis data would not provide useful information be-cause the loops would close during the rocking of the joint. However, cycling the column through a series of reversible loads would provide information on the strength of the column, the ability of the joint to transfer moment, and the relative stiffness of the connection. Figure 11 gives the ap-plied cyclic loading pattern for testing. The lateral load at the top of the column was applied by alternately stressing

Conclusions from the column-capital test

The capital section was stiff, with less than 0.3 in. (7.6 mm) of vertical deflection under service load. This de-flection includes the effects of rotation by the column-cap-ital and residual torsional effects. The capital failure was ductile, with warning of imminent failure demonstrated by ample deflection and cracking. The shear capacity exceed-ed the flexural capacity, demonstrating that the details used in the specimen were adequate.

ACI 318-05 Eq. (11-5) has a maximum limit of concrete stress of 3.5 fc

' psi (0.3 fc' MPa), which was the limit-

ing condition for this specimen. Designing to this higher limit was conservative for this application. However, the flexure failure precluded knowing the exact shear capacity. A design based on 2 fc

' psi (0.33 fc' MPa) plus the stir-

rup contribution may be more appropriate for final design if longer shear spans are used. Punching shear was not a

Figure 11. This cyclic sequence was used to develop the column lateral loading. Note: 1 in. = 2.54 mm.

-3.00

-2.00

-1.00

0.00

1.00

2.00

3.00

0 10 20 30 40 50 60 70 80

Top

dis

plac

emen

t, in

.

Number of cycles

Figure 12. This is an illustration of the cyclic load-deflection history of the column connection. Note: 1 in. = 2.54 mm; 1 kip = 4.45 kN.

-3000 -2000

-1000 0

1000 2000 3000 4000

-3.00 -2.00 -1.00 0.00 1.00 2.00 3.00

Bas

e m

omen

t, ki

p-in

.

Top displacement, in.

West East


the two hydraulic cylinders and the threaded bar/pipe load-ing mechanism shown on top of the column in Fig. 4.

Figure 12 shows the loads corresponding to the deflection at the end of each cycle. Only the final load for each cycle was recorded except for the final east push, for which several intermediate data points were recorded. Figure 12 indicates that the inability to tighten the connection af-fected the relative stiffness.

The push in the west direction resulted in deflections nearly twice those of the east direction for comparable loads. This reflected the rocking that closed the joint, followed by an increase in load once the joint tightened. The column rein-forcement did not yield in the west direction. The column steel yielded at about 2.2% drift, or 2 in. (50 mm) deflec-tion, in the east direction. The final push in the east direction showed that the initial column movement was followed by joint tightening at about 1.0 in. (25 mm) horizontal deflec-tion. The column reinforcement yielded when the load was increased to produce a 1.4 in. (35 mm) deflection.

Following yield, the reaction became nonlinear. The test was stopped at 3 in. (150 mm) top deflection, or about 3.3% drift. At 3.3% drift, the base moment was 3026 kip-in. (339 kN-m) versus a predicted capacity of 2700 kip-in. (303 kN-m) with no axial load. The moment at the level of the connection was 2220 kip-in. (249 kN-m), and damage to the column face was evident. The test developed a bending load that exceeded the theoretical nominal capacity of the column. The dead load of the col-umn was not included in the calculations and would have a slight effect in increasing the predicted moment capacity.

Figure 13. The column crack pattern corresponds to a 2 in. (50 mm) top displace-ment.

Figure 14. A ½ in. lateral slip in the column-to-column joint occurred at a top deflection of 1.75 in. Note: 1 in. = 25.4 mm.


stiffness for erection. Shear restraint was included to limit the lateral movement in the joint.

Final construction details

The beam slab, rib slab, hanger, and column capital were implemented as tested. The column connection was modi-fied by partially developing four no. 11 (36M) mild-steel reinforcing bars and using bolted connections located at middepth of the column face (Fig. 15). The connection was designed to provide sufficient stiffness and strength for unbalanced erection and construction loads without the need for bracing on interior columns. In addition to adequate moment capacity, the revised detail provided a superior concrete detail at the column corner. The change in column details did not affect the column capital performance.

Acknowledgment

This research was sponsored by Rocky Mountain Pre-stress. A patent application has been submitted for the building system described in this paper.

An assessment of the cracks in the column showed a uniformly distributed crack pattern above and below the connection on the west face of the column (Fig. 13). Even with the loose connectors, the moment connection was fully developed on both sides of the joint.

Damage was also visible around the connection blockouts. The damage was caused by lateral strain in the connectors and the stress concentration due to connectors bearing on the steel plates at the bottom of the blockout.

Once the top deflection reached 1.75 in. (44 mm), a lateral slip was observed. The slip varied from about 1/4 in. (6 mm) at a top deflection of 1.75 in. (44 mm) to nearly 7/8 in. (22 mm) at 3 in. (150 mm) of top displacement. Figure 14 shows the lateral slip at the third cycle at a top deflection of 1.75 in. (44 mm). The production design incorporates an internal shear key to control the lateral slip.

Column test conclusions and recommendations

The column test successfully demonstrated that the nomi-nal capacity of the section could be developed and that the connection successfully transferred the moment to the stub section below. The tolerance of the connection required resolution prior to construction to provide sufficient lateral

Figure 15. The final field column-to-column connection places the connection at the midface of the column and no. 11 (36M) mild-steel reinforcing bars welded to the ¾-in.-thick (19 mm) steel plate.


References

1. Priestly, M. J. N. 1991. Overview of PRESSS Re-search Program. PCI Journal, V. 36, No. 4 (July–August): pp. 50–57.

2. PCI Industry Handbook Committee. 2004. PCI De-sign Handbook: Precast and Prestressed Concrete. 6th ed. Chicago, IL: PCI.

3. American Concrete Institute (ACI) Committee 318. 2005. Building Code Requirements for Structural Concrete (ACI 318-05) and Commentary (ACI 318R-05). Farmington Hills, MI: ACI.

4. ACI Innovative Task Group 1.1. 2003. Special Hy-brid Moment Frames Composed of Discretely Jointed Precast and Post-tensioned Concrete Members (T1.2-03) and Commentary (T1.2R-03). Farmington Hills, MI: ACI.

Notation

Av = area of shear reinforcement

b = width of compression face of member

bc = cross-sectional dimension of column core measured center to center of outer legs of the transverse reinforcement

d = distance from extreme compression fiber to centroid of longitudinal tension reinforcement

fc' = specified compressive strength of concrete

s = center-to-center spacing of shear reinforcement

Vc = nominal shear strength provided by the concrete

Vn = nominal shear strength


About the authors

John Hanlon is vice president of engineering at Rocky Mountain Prestress in Denver, Colo., and is the innovator of the precast concrete building system presented in this research.

Charles W. Dolan, FPCI, is the H. T. Person Professor of Engineer-ing at the University of Wyoming in Laramie, Wyo., and served as chair of subcommittee G, Precast and Prestressed Concrete, of ACI Committee 318, Building Code for Concrete Structures.

David Figurski is production supervisor at Rocky Mountain Prestress in Denver. He received his master’s degree at the University of Texas in alkali-silica reactions in concrete and has been at Rocky Mountain Prestress for seven years.

Jiangang Deng is a graduate research assistant at the Univer-sity of Wyoming, where he is researching the durability of epoxy-bonded, fiber-reinforced-polymer-strengthened structures.

J. Gregory Dolan was an under-graduate research assistant at the University of Wyoming.

Synopsis

This research and development effort examines the design and performance of elements required for successful construction of a precast concrete build-ing system that emulates flat-slab construction. The research focused on the development and testing of the rib slab, the column connection, and the precast concrete column capital.

The research confirmed the flexural and connec-tion capacity of the rib slab, the flexural and shear capacity of the precast concrete column capital, and the moment-resisting capacity of the precast concrete column connection. The test program identified a lack of stiffness with the column-to-column connection and developed solutions for this connection.

Keywords

Building, capital, column, connection, emulative de-sign, flat slab, hanger, rib slab, system, testing.

Review policy

This paper was reviewed in accordance with the Precast/Prestressed Concrete Institute’s peer-review process.

Reader comments

Please address any reader comments to PCI Journal editor-in-chief Emily Lorenz at [email protected] or Precast/Prestressed Concrete Institute, c/o PCI Journal, 209 W. Jackson Blvd., Suite 500, Chicago, IL 60606. J

pc frame building components for westin hotel part 1

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