THE UNIVERSITY OF HONG KONGLIBRARIES

Hong Kong Collectiongift from

Hong Kong. Civil Engineering Dept.

GEO PUBLICATION No. 1/96

PILE DESIGNAND CONSTRUCTION

GEOTECHNICAL ENGINEERING OFFICECivil Engineering DepartmentHong Kong

*%

© Hong Kong Government

First published, June 1996

Prepared by:

Geotechnical Engineering Office,Civil Engineering Department,Civil Engineering Building,101 Princess Margaret Road,Homantin, Kowloon,Hong Kong.

Captions of Figures on the Front Cover :

Top Left Construction of Large-diameter Bored Piles on Slope

Top Right Distribution of Compressive Strain and Force in Pile

Bottom Left Foundation in Karst Marble

Bottom Right Construction of Large-diameter Bored Piles in the Sea

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FOREWORD

This publication presents a review of the principles and practice related to the designand construction of piles, with specific reference to ground conditions in Hong Kong. Theinformation given herein should facilitate the use of modern methods and knowledge in thedesign of piles.

Given the geology of the main urban areas of Hong Kong, where sandy soils derivedfrom igneous rocks by weathering predominate, emphasis is placed on the design of piles ingranular soils. The approaches to the design of piles in clayey soils, rock and karstconditions are also presented.

The process of constructing or installing a pile can alter the engineering properties ofthe ground and different construction techniques may produce quite different pileperformance. Consideration of construction control should therefore be an integral part ofpile design. Guidance is provided on construction control, and problems that may beencountered during construction are discussed, along with the possible ways to avoid theseproblems or mitigate their effects.

The data available to us from instrumented pile load tests in Hong Kong are collatedin this document. Practitioners are invited to help expand this database by continuing toprovide us with raw data from local instrumented pile load tests.

This document was prepared initially by a team consisting of Mr S.H. Tse, Mr S.H.Mak, Mr K.K.S. Ho, Dr T.S.K. Lam, Mr C.K. Wong and Mr T.C.F. Chan. The presentedition was written by Mr K.K.S. Ho and finalised by Dr D.O.K. Lo, under the supervisionof Mr S.H. Mak initially and later Mr W.K. Pun. Mr Y.C. Chan directed the project from1991 to 1996.

Copies of a draft version of this document were circulated locally and abroad forcomment in 1995. Those consulted included consulting engineers, contractors, academics,professional bodies and Government departments. Many individuals and organisations madevery valuable comments, many of which were adopted in finalising this document, and theircontributions are gratefully acknowledged. These comments, together with the responses,were made up into a volume which has been lodged in the Civil Engineering Library whereit may be inspected by interested parties.

Practitioners are encouraged to comment to the Geotechnical Engineering Office onthe contents of this publication, so that improvements can be made in future editions.

A.W. MalonePrincipal Government Geotechnical Engineer

June 1996

CONTENTS

PageNo.

TITLE PAGE l

FOREWORD 3

CONTENTS 5

1. INTRODUCTION 13

1.1 PURPOSE AND SCOPE 13

1.2 GENERAL GUIDANCE 14

2. SITE INVESTIGATION, GEOLOGICAL MODELS AND 15SELECTION OF DESIGN PARAMETERS

2.1 GENERAL 15

2.2 DESK STUDIES 152.2.1 Site History 152.2.2 Details of Adjacent Structures and Existing Foundations 162.2.3 Geological Studies 162.2.4 Groundwater 17

2.3 EXECUTION OF GROUND INVESTIGATION 18

2.4 EXTENT OF GROUND INVESTIGATION 182.4.1 General 182.4.2 Sites Underlain by Marble 19

2.5 SOIL AND ROCK SAMPLING 20

2.6 DETECTION OF AGGRESSIVE GROUND 20

2.7 INSITU AND LABORATORY TESTING 21

2.8 ESTABLISHMENT OF GEOLOGICAL MODEL 22

PageNo.

2.9 SELECTION OF DESIGN PARAMETERS 23

3. TYPES OF PILE 25

3.1 CLASSIFICATION OF PILES 25

3.2 LARGE-DISPLACEMENT PILES 263.2.1 General 263.2.2 Precast Reinforced Concrete Piles 263.2.3 Prestressed Concrete Tubular Piles 263.2.4 Closed-ended Steel Tubular Piles 273.2.5 Driven Cast-in-place Concrete Piles 27

3.3 SMALL-DISPLACEMENT PILES 273.3.1 General 273.3.2 Steel H-piles 283.3.3 Open-ended Steel Tubular Piles 28

3.4 REPLACEMENT PILES 283.4.1 General 283.4.2 Machine-dug Piles 283.4.3 Hand-dug Caissons 29

3.5 SPECIAL PILE TYPES 323.5.1 General 323.5.2 Barrettes 323.5.3 Mini-piles 333.5.4 Composite Piles 33

4. CHOICE OF PILE TYPE AND DESIGN RESPONSIBILITY 35

4.1 GENERAL 35

4.2 FACTORS TO BE CONSIDERED IN CHOICE OF PILE TYPE 354.2.1 General 354.2.2 Ground Conditions 354.2.3 Nature of Loading 374.2.4 Effects of Construction on Surrounding 37

Structures and Environment4.2.5 Site and Plant Constraints 384.2.6 Safety 384.2.7 Programme and Cost 39

PageNo.

4.3 DESIGN RESPONSIBILITY 394.3.1 Contractor's Design 394.3.2 Engineer's Design 404.3.3 Discussion 40

5. DESIGN OF SINGLE PILES AND DEFORMATION OF PILES 43

5.1 GENERAL 43

5.2 PILE DESIGN IN RELATION TO GEOLOGY 43

5.3 DESIGN PHILOSOPHIES 445.3.1 General 445.3.2 Global Factor of Safety Approach 445.3.3 Partial Factor of Safety Approach 445.3.4 Discussion 455.3.5 Recommended Factors of Safety 46

5.4 AXIALLY LOADED PILES IN SOIL 475.4.1 General 475.4.2 Pile Driving Formulae 475.4.3 Wave Equation Analysis 495.4.4 Use of Soil Mechanics Principles 505.4.5 Correlation with Standard Penetration Tests 555.4.6 Correlation with Other Insitu Tests 57

5.5 AXIALLY LOADED PILES IN ROCK 585.5.1 General 585.5.2 Driven Piles on Rock 585.5.3 Bored Piles on Rock 595.5.4 Rock Sockets 61

5.6 UPLIFT CAPACITY OF PILES 625.6.1 Piles in Soil 625.6.2 Rock Sockets 635.6.3 Cyclic Loading 64

5.7 LATERAL LOAD CAPACITY OF PILES 645.7.1 Vertical Piles in Soil 645.7.2 Inclined Loads 665.7.3 Raking Piles in Soil 665.7.4 Rock Sockets 665.7.5 Cyclic Loading 67

PageNo.

5.8 NEGATIVE SKIN FRICTION 675.8.1 General 675.8.2 Calculation of Negative Skin Friction 675.8.3 Field Observations in Hong Kong 695.8.4 Means of Reducing Negative Skin Friction 70

5.9 TORSION 70

5.10 PRELIMINARY PILES FOR DESIGN EVALUATION 70

5.11 PILE DESIGN IN KARST MARBLE 72

5.12 STRUCTURAL DESIGN OF PILES 735.12.1 General 735.12.2 Lifting Stresses 735.12.3 Driving and Working Stresses 735.12.4 Bending and Buckling of Piles 745.12.5 Mini-piles 74

5.13 DEFORMATION OF SINGLE PILES 755.13.1 General 755.13.2 Axial Loading 755.13.3 Lateral Loading 79

5.14 CORROSION OF PILES 83

GROUP EFFECTS 87

6.1 GENERAL 87

6.2 MINIMUM SPACING OF PILES 87

6.3 ULTIMATE CAPACITY OF PILE GROUPS 886.3.1 General 886.3.2 Vertical Pile Group in Granular Soils under Compression 896.3.3 Vertical Pile Group in Clays under Compression 906.3.4 Vertical Pile Group in Rock under Compression 916.3.5 Vertical Pile Group under Lateral Loading 916.3.6 Vertical Pile Group under Tension Loading 926.3.7 Pile Groups Subject to Eccentric Loading 92

6.4 NEGATIVE SKIN FRICTION ON PILE GROUPS 93

PageNo.

6.5 DEFORMATION OF PILE GROUPS 946.5.1 Axial Loading on Vertical Pile Group 946.5.2 Lateral Loading on Vertical Pile Group 986.5.3 Combined Loading on General Pile Group 99

6.6 DESIGN CONSIDERATIONS IN SOIL-STRUCTURE 100INTERACTION PROBLEMS6.6.1 General 1006.6.2 Load Distribution between Piles 1016.6.3 Behaviour of Piled Rafts at Working Condition 1036.6.4 Use of Piles to Control Foundation Stiffness 1046.6.5 Piles in Soils Undergoing Movement 105

7. PILE INSTALLATION AND CONSTRUCTION CONTROL 107

7.1 GENERAL 107

7.2 INSTALLATION OF DISPLACEMENT PILES 1077.2.1 Equipment 1077.2.2 Characteristics of Hammers and Vibratory Drivers 1087.2.3 Selection of Method of Pile Installation 1107.2.4 Potential Problems Prior to Pile Installation 1117.2.5 Potential Problems during Pile Installation 1127.2.6 Potentially Damaging Effects of Construction and 121

Mitigating Measures

7.3 INSTALLATION OF MACHINE-DUG PILES 1257.3.1 Equipment 1257.3.2 Use of Drilling Fluid for Support of Excavation 1267.3.3 Assessment of Founding Level and Condition of Pile Base 1277.3.4 Potential Problems during Pile Excavation 1297.3.5 Potential Problems during Concreting 1337.3.6 Potential Problems after Concreting 138

7.4 INSTALLATION OF HAND-DUG CAISSONS 1397.4.1 General 1397.4.2 Assessment of Condition of Pile Base 1397.4.3 Potential Installation Problems and Construction 140

Control Measures

7.5 INTEGRITY TESTING OF PILES 1437.5.1 Role of Integrity Testing 1437.5.2 Types of Non-destructive Integrity Testing 144

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PageNo.

7.5.3 Practical Considerations in the Use of Integrity Tests 150

8. PILE LOAD TESTS 153

8.1 GENERAL 153

8.2 TIMING OF PILE TESTS 153

8.3 STATIC LOAD TESTS 1548.3.1 Reaction Arrangement 1548.3.2 Equipment 1558.3.3 Test Procedures 1568.3.4 Instrumentation 1588.3.5 Interpretation of Test Results 160

8.4 DYNAMIC LOAD TESTS 1638.4.1 General 1638.4.2 Test Methods 1648.4.3 Methods of Interpretation 1648.4.4 Recommendations on the Use of Dynamic Load Tests 166

REFERENCES 169

TABLES 203

LIST OF TABLES 205

TABLES 207

FIGURES 237

LIST OF FIGURES 239

FIGURES 243

APPENDIX A SUMMARY OF RESULTS OF INSTRUMENTED 303PILE LOAD TESTS IN HONG KONG

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PageNo..

GLOSSARY OF SYMBOLS 331

GLOSSARY OF TERMS 345

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1. INTRODUCTION

1.1 PURPOSE AND SCOPE

The purpose of this document is to give guidance for the design and construction ofpiles in Hong Kong. It is aimed at professionals and supervisory personnel involved in thedesign and construction of piles. The document has been prepared on the assumption that thereader has some general knowledge of piling.

In the context of this document, a pile refers to a columnar foundation element thatis :

(a) pre-manufactured and inserted into the ground by driving,jacking or other methods, or

(b) cast-in-place in a shaft formed in the ground by boring orexcavation.

Traditional pile design practice in Hong Kong relies, in part, on the British Code ofPractice for Foundations (BSI, 1954), together with empirical rules formulated some 30 yearsago from local experience with piling in weathered granite profiles. The enactment of therevised Building (Construction) Regulations (Hong Kong Government, 1990) marks a turningpoint in local foundation practice in that rational design approaches based on acceptedengineering principles are now formally recognised. This requires a greater geotechnicalinput including properly-planned site investigations, field and laboratory testing, together withconsideration of the method of construction. The use of rational methods to back-analyseresults of load tests on instrumented piles or the monitored behaviour of prototype structureswill lead to a better understanding of pile behaviour. This should enable more reliable andeconomical design to be employed in future projects.

A thorough understanding of the ground conditions is a pre-requisite to the success ofa foundation project. An outline of geological conditions pertinent to piling is given inChapter 2, along with guidance on the scope of site investigations required for the design ofpiles.

In Hong Kong, tall buildings in excess of 30 storeys are commonplace both onreclamations and on hill slopes. Steel and concrete piles are generally used as buildingfoundations. Timber piles, which were used extensively in the past to support low-risebuildings and for wharves and jetties, are not covered in this document. Guidance on thetypes of piles commonly used in Hong Kong is given in Chapter 3.

Factors to be considered in choosing the most appropriate pile type and the issue ofdesign responsibility are discussed in Chapter 4.

Guidance on methods of designing single piles and methods of assessing pile movementare given in Chapter 5. The design of pile groups and their movement are covered inChapter 6. Given the nature of the geology of the urban areas of Hong Kong where granularsoils predominate, emphasis has been placed on the design of piles in granular soil and

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weathered rock, although pile design in clay has also been outlined for use in areas underlainby argillaceous rock.

Consideration of the practicalities of pile installation and the range of constructioncontrol measures form an integral part of pile design, since the method of construction canhave a profound influence on the ground and hence on pile performance. A summary of pileconstruction techniques commonly used in Hong Kong and a discussion on the range ofpotential construction problems, together with possible precautionary measures that may beadopted, are given in Chapter 7.

In view of the many uncertainties inherent in the design of piles, it is difficult topredict with accuracy the behaviour of a pile, even with the use of sophisticated analyses.The actual performance of single piles is best verified by a load test, and foundationperformance by building settlement monitoring. Chapter 8 describes the types of, andprocedures for, static and dynamic load tests commonly used in Hong Kong.

1.2 GENERAL GUIDANCE

In this document, reference has been made to published codes, textbooks and otherrelevant information. The reader is strongly advised to consult the original publications forfull details of any particular subject.

The various stages of site investigation, design and construction of piles require a co-ordinated input from experienced personnel. Pile design is not complete upon the productionof construction drawings. Continual involvement of the designer is essential in checking thevalidity of both the geological model and the design assumptions as construction proceeds.As the installation method may significantly affect pile performance, it is most important thatexperienced and competent specialist contractors are employed and their work adequatelysupervised by suitably qualified and experienced engineers who should be knowledgeableabout the basis of the design.

In common with other types of geotechnical structures, professional judgement andengineering common sense must be exercised when designing and constructing piles.

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2. SITE INVESTIGATION, GEOLOGICAL MODELS ANDSELECTION OF DESIGN PARAMETERS

2.1 GENERAL

A thorough understanding on the ground conditions of a site is a pre-requisite to thesuccess of a foundation project. The overall objective of a site investigation for foundationdesign is to determine the site constraints, geological profile and the properties of the variousstrata. The geological sequence can be established by sinking boreholes from which soil androck samples are retrieved for identification and testing. Insitu tests may also be carried outto determine the mass properties of the ground. These investigation methods may besupplemented by regional geological studies and geophysical tests where justified by the scaleand importance of the project, or the complexity of the ground conditions.

The importance of a properly planned and executed ground investigation cannot beover-emphasized. The information obtained from the investigation will allow an appropriategeological model to be constructed. This determines the selection of the optimum foundationsystem for the proposed structure. It is important that the engineer planning the siteinvestigation and designing the foundations liaises closely with the designer of thesuperstructure and the project coordinator so that specific requirements and site constraintsare fully understood by the project team.

An oversimplified ground investigation is a false economy as it can lead to delaysduring construction and substantial cost overruns. The investigation should always beregarded as a continuing process that requires regular re-appraisals. For large projects orsites with a complex geology, the investigation should be phased to allow appropriateamendments of the study schedule in response to the actual sub-surface conditionsencountered. In some cases, additional ground investigation may be necessary during, orsubsequent to, pile construction. For maximum cost-effectiveness, it is important to ensurethat appropriate tests are undertaken to derive relevant design parameters.

General guidance on the range of site investigation methods is given in Geoguide 2 :Guide to Site Investigation (GCO, 1987), which is not repeated here. Specific guidancepertinent to marine investigations is given in BS 6349 (BSI, 1984). This Chapter highlightsthe more important aspects of site investigation with respect to pile foundations.

2.2 DESK STUDIES

2.2.1 Site History

Information on site history can be obtained from various sources including plans ofprevious and existing developments, aerial photographs, old topographic maps, together withgeological maps and memoirs. Useful information on the possible presence of oldfoundations, abandoned wells, tunnels, etc., may be extracted from a study of the site history.For sites on reclaimed land or within areas of earthworks involving placement of fill, it isimportant to establish the timing and extent of the reclamation or the earthworks, based onaerial photographs or old topographic maps, to help to assess the likelihood of continuing

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ground settlement that may give rise to negative skin friction on piles. Morrison & Pugh(1990) described an example of the use of this information in the design of foundations. Oldpiles left behind in the ground from demolition of buildings may affect the design andinstallation of new piles. It is important to consider such constraints in the choice of pile typeand in designing the pile layout.

Sites with a history of industrial developments involving substances which maycontaminate the ground (e.g. dye factories, oil terminals) will require detailed chemicaltesting to evaluate the type, extent and degree of possible contamination.

2.2.2 Details of Adjacent Structures and Existing Foundations

Due to the high intensity of developments in Hong Kong, a detailed knowledge ofexisting structures and their foundations, including tunnels, within and immediately beyondthe site boundaries is important because these may pose constraints to the proposed foundationconstruction. Records and plans are available in the Buildings Department for privatedevelopments, and in the relevant government offices for public works. Details of theexisting foundation types and their construction and performance records will serve as areference for the selection of the most appropriate foundation type for the proposeddevelopment. In certain circumstances, it may be feasible or necessary to re-use some of theexisting foundations if detailed records are available and their integrity and capacity can beconfirmed by testing.

Particular attention should be paid to the special requirements for working in thevicinity of the Mass Transit Railway (MTR) (designated as Scheduled Area No.3), provisionsfor Scheduled Area Nos. 1 (Mid-levels), 2 (Yuen Long) and 4 (Ma On Shan) as stipulatedby Hong Kong Government (1993), possible presence of sensitive apparatus (e.g. computers,specialist machinery, etc.) within adjacent buildings, and locations of hospitals or otherbuildings having special purposes that may have specific requirements. Attention should alsobe paid to the deep sewage tunnels, and other existing tunnels, caverns, service reservoirs andrailways. All these may pose constraints on the construction work.

2.2.3 Geological Studies

An understanding of the geology of the site is a fundamental requirement in theplanning and interpretation of the ground investigation. The solid and superficial geology ofHong Kong is varied. A useful summary of the nature and occurrence of rocks and soils inHong Kong can be found in Geoguide 3 : Guide to Rock and Soil Description (GCO, 1988).Detailed information about geology of Hong Kong can be obtained from the latest maps andmemoirs, provided at several scales, by the Hong Kong Geological Survey. A generalisedgeological map of Hong Kong is shown in Figure 1. The broad divisions of the principalrock and soil types are summarised in Table 1. Given the diverse geological conditions, itis inadvisable to follow any pile design rules uniformly without regard to geologicalvariations.

Typically, a mantle of weathered rock, and colluvium on hillsides, overlies fresh rock.The thickness and nature of the weathering profiles vary markedly, depending on rock type,

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geological history and topographical location. Corestone-bearing profiles (Figure 2),primarily developed in the medium- and coarse-grained granites and coarse ash tuffs (volcanicrock), are common on hillslopes although exposures have also indicated that many volcanicrocks and some granites do not contain corestones, and the transition from soil to rock isgradual. The incidence of corestones generally increases with depth in a weathering profile.This, coupled with the inherent spatial variability of the soil masses formed from weatheringof rocks insitu and the undulating rock surface, constitute important considerations in thedesign and construction of piles in Hong Kong.

Most of the piling experience in Hong Kong has been derived from foundations in theweathered profile in the granitic formations. Granitic saprolites (i.e. mass which retain theoriginal texture, fabric and structure of the parent rock) are generally regarded as granularsoils in terms of their engineering behaviour. In addition, they may possess relict orsecondary bonding, depending on the degree of weathering and cementation.

Published accounts of piling in volcanic saprolite are limited. The lithologicalvariability of volcanic rocks can be considerable; they occur as recrystallised and weldedtuffs, lava flows, and may be metamorphosed; some are sedimentary rocks of volcanic originranging in grain size from mudstones to conglomerates. The rate and products of weatheringof these rocks vary widely, most soils derived from volcanic rocks may be silty. They maycontain fragile, partially or wholly decomposed grains and possess relict bonding. In viewof the diversity of rock types, structural geology and weathering profiles, generalisation aboutpiling in volcanic rocks is inadvisable.

Colluvium including debris flow and rockfall deposits, have accumulated commonlyon the hillsides, and may extend to the lower reaches of valleys. Numerous boulders maybe present within a generally medium-grained to coarse-grained matrix, which may impedepile driving. Clay profiles are generally rare in weathered rock. However, clays may bepresent as alluvial clays or the fine-grained material of weathered meta-siltstones of the LokMa Chau Formation (Table 1).

For sites underlain by karst marble, the foundations may be adversely affected bydissolution voids (GCO, 1990). Stability of the foundation will depend on the particulargeometry of the dissolution features and the rock mass properties.

It is important to note the significance of careful field observations and experience inrelation to the influence of geology on pile performance. Such an experience database, builton a direct and empirical relationship between geology and engineering, can be invaluable,particularly in circumstances where observations cannot be adequately explained by the theoryof mechanics. On the other hand, it must be cautioned that experience can becomegeneralized as rules of thumb. It is advisable to be aware of the danger of these rales ofthumb being invalidated by differences in geology, or variations in mechanical behaviour ofthe range of materials in a given geological formation.

2.2.4 Groundwater

Information on the groundwater regime is necessary for the design and selection of piletype and method of construction. Artesian water pressures may adversely affect shaft stability

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for cast-in-place piles. For developments close to the seafront, the range of tidal variationsshould be determined. In a sloping terrain, there may be significant groundwater flow, andhence the hydraulic gradients should be determined as far as possible since the flow can affectthe construction of cast-in-place piles, and the consideration of possible damming effects mayinfluence the pile layout in terms of the spacing of the piles.

2.3 EXECUTION OF GROUND INVESTIGATION

It is essential that experienced and competent ground investigation contractors with aproven track record and capable of producing high quality work are employed in groundinvestigations. The fieldwork should be designed, directed and supervised by a qualified andexperienced engineer or engineering geologist, assisted by trained and experienced technicalpersonnel where appropriate. Suitable levels of supervision of ground investigation worksare recommended in Geoguide 2 : Guide to Site Investigation (GCO, 1987).

2.4 EXTENT OF GROUND INVESTIGATION

2.4.1 General

The extent of a ground investigation is dependent on the complexity of the ground and,to a certain degree, the form of the proposed development and the intended pile types.Adequate investigation should be carried out to ensure no particular piling options will beprecluded due to a lack of information on ground conditions. Although no hard and fast rulescan be laid down, a relatively close borehole spacing of say 10 m to 30 m will often beappropriate for general building structures. In reclamation areas, closely-spaced boreholesmay be needed to delineate buried obstructions such as remnants of an old seawall where thisis suspected from a desk study of the site history.

In general, boreholes should be extended through unsuitable founding materials intocompetent ground beyond the zone of influence of the proposed development. Where pilefoundations are considered to be a possibility, the length of pile required usually cannot bedetermined until an advanced stage of the project. Some general guidance in this instance isgiven in Geoguide 2 : Guide to Site Investigation (GCO, 1987).

The traditional ground investigation practice in Hong Kong is to sink boreholes to atleast 5 m into Grade III or better rock to prove that a boulder has not been encountered. Thispractice should not be regarded as reliable unless backed by a geological model prepared bya suitably experienced professional.

It is good practice to sink sufficient boreholes to confirm the general geology of thesite. Consideration should also be given to sinking boreholes immediately outside the loadedarea of a development in order to improve the geological model. It is also important tocontinually review the borehole findings throughout the investigation stage to ensure adequateinformation has been obtained.

For large developments or sites with a complex geology, it is advisable to phase theinvestigation to enable a preliminary geological assessment and optimal planning of the later

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stages of investigation. Where large-diameter bored piles have been chosen, the usualpractice in Hong Kong is to sink one borehole at each pile position (or sufficient boreholesfor individual groups of closely-spaced piles as appropriate), during the construction stage andprior to pile installation, to a depth of three times the pile diameter below the intendedfounding level in order to confirm the design assumptions. The above approach should alwaysbe adopted in Hong Kong in view of the inherent variability of ground conditions and thepossible presence of corestones in the weathering profile.

Where appropriate, geophysical methods may be used to augment boreholes. A rangeof surface, cross-hole and down-hole geophysical techniques (Braithwaite & Cole, 1986;GCO, 1987) are available. The undertaking and interpretation of geophysical surveys requirea sound knowledge of the applicability and limitations of the different techniques, properunderstanding of geological processes and the use of properly calibrated equipment. The datashould be processed in the field as far as possible in order that apparent anomalies may beresolved or confirmed. Geophysical techniques are generally useful in helping to screen thesite area for planning of the subsequent phases of investigation by drilling.

2.4.2 Sites Underlain by Marble

Given the possible extreme variability in karst morphology of the marble rock mass,the programme of ground investigation should be flexible. It is important that the boreholelogs and cores are continuously reviewed as the work progresses so that the investigationworks can be suitably modified to elucidate any new karst features intercepted.

For high-rise developments on sites underlain by marble, the investigation should bestaged and should be carried out under the full-time supervision of qualified personnel. Forpreliminary investigation, it is recommended that there should be a minimum of one boreholeper 250 m2, drilled at least 20 m into sound marble rock, i.e. rock which has not beenaffected, or is only slightly affected, by cavities (Chan, 1994b). The position of subsequentboreholes for determining the extent of dissolution features, such as overhanging pinnacles,deep cavities and floaters, should be based on the findings of the preliminary boreholes. Itis anticipated that boreholes will be required to intercept specific karst features on a grid ofabout 7 m to 10 m centres. Boreholes in other parts of the site should be sunk on a gridpattern or at points of concentration of piles, to a depth of 20 m into sound marble. Attentionshould be given to logging the location and size of cavities, the nature of the cavity walls,infilling materials and discontinuities. If the infill is cohesive in nature, good quality tubesamples of cavity infill may be obtained using triple-tube core barrels with water, orpreferably air foam, as the flushing medium.

A lower intensity of borehole spacing may be sufficient for low-rise developments.Where the loading is small or where the superficial deposits above the marble rock are verythick, drilling may be limited to a depth where there is a minimum of 20 m of competentfounding material. Nevertheless, it is strongly recommended that at least one deep boreholeis sunk at each site underlain by marble, say to 100 m below ground level, to confirm thegeological profile.

Surface geophysical methods are generally ineffective for the investigation of marblerock conditions because the marble in Hong Kong is generally overlain by 30 m to 70 m of

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superficial deposits (Langford, 1990). Borehole geophysical techniques, including cross-holeseismic shooting and electro-magnetic wave logging, have been found to give meaningfulresults. However, geophysical measurements should only be regarded as supplementaryground investigation methods in view of their inherent limitations and the simplificationsinvolved in the interpretation. The value of geophysical testing is that it gives a greater levelof confidence in the adequacy of the ground investigation, particularly in relation to theground conditions existing between adjacent boreholes. In addition, the results may be usedto help positioning the boreholes of the subsequent phase of ground investigation.

All boreholes must be properly grouted upon completion of drilling. This is especiallyimportant in the case of drilling into cavernous marble in order to minimise the risk of groundloss and sinkhole formation arising from any significant water flow that may otherwise bepromoted.

2.5 SOIL AND ROCK SAMPLING

Wash boring with no sampling is strongly discouraged. It is recommended practicealways to retrieve good quality soil samples and continuous rock cores from boreholes forboth geological logging and laboratory testing. A possible exception to this can be made forsupplementary boreholes sunk solely for the purposes of investigating particular dissolutionfeatures in cavernous marble.

Good quality samples of soils derived from insitu rock weathering can be retrievedusing triple tube core barrel (e.g. Mazier samplers). Detailed logging of the geologicalprofile using such soil samples can help to identify salient geological features.

2.6 DETECTION OF AGGRESSIVE GROUND

In general, materials derived from the insitu weathering of rocks in Hong Kong arenot particularly aggressive to concrete and steel, unless they have been contaminated.However, marine muds, estuarine deposits and fill can contain sulphate-reducing bacteria orother deleterious constituents that may pose a potential risk of damaging the pile material.In reclaimed land, the content of sulphate or other corrosive trace elements may be up tolevels that give cause for concern. The zone within the tidal or seasonal water tablefluctuation range is generally most prone to corrosion because of more intensive oxidation.In industrial areas or landfill sites, the waste or contaminated ground may impede setting ofconcrete or attack the pile material.

Basic chemical tests on soil and groundwater samples including the determination ofpH and sulphate content (total and soluble) should be carried out where necessary. For sitesclose to the seafront, the saline concentration of groundwater should be determined. In sitesinvolving landfills or which are close to landfills, the possible existence of toxic leachate orcombustible gases (such as methane) or both, and the rates of emission should be investigated,paying due regard to the possibility of lateral migration. Enough information should becollected to assess the risk of triggering an underground fire or a surface explosion duringfoundation construction (e.g. during welding of pile sections) in such sites.

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Where other deleterious chemicals are suspected (e.g. on the basis of site history),specialist advice should be sought and relevant chemical tests specified. For instance, heavymetal contamination (especially lead and mercury) can, depending on the degree of solubilityor mobility in water, represent a health risk to site workers. The degree of contaminationcan dictate the means by which the spoil from pile excavation will have to be disposed of.It should also be noted that high levels of organic compounds including oils, tars and greases(as reflected by, for instance, toluene extractable matter measurements) can severely retardor even prevent the setting of concrete, or alternatively can potentially cause chemical attackof concrete at a later stage (Section 5.14). It should be noted that particular safetyprecautions should be taken when investigating a landfill or contaminated site.

Various classification systems have been proposed to assess the degree ofcontamination of a site, e.g. Kelly (1980) and Department of the Environment (1987).

2.7 INSITU AND LABORATORY TESTING

For a rational pile design, it is necessary to have data on the strength andcompressibility of the soil and rock at the appropriate stress levels within the zone ofinfluence of the proposed development. Other relevant parameters include permeability, suchas for foundation works involving dewatering or grouting, and the properties of rock jointsfor the design of a laterally-loaded rock socket.

Insitu tests are usually carried out during the ground investigation. The range ofcommonly-used tests includes Standard Penetration Test (SPT), Cone Penetration Test (CPT)and piezocone, pressuremeter, plate bearing, vane shear, insitu permeability, impressionpacker and light weight probes.

It should be noted that the state and properties of the ground may change as a resultof pile installation. Where deemed appropriate, test driving or trial bore construction maybe considered as an investigative tool to prove the feasibility of construction methods and theadequacy of quality control procedures.

Laboratory testing should be carried out to complement information obtained frominsitu tests to help to characterise the material and determine the relevant design parameters.The tests may be grouped into two general classes :

(a) Classification or index tests - for grouping soils with similarengineering properties, e.g. particle size distribution,Atterberg Limits, moisture content, specific gravity,petrographic examination.

(b) Quantitative tests - for measurement of strength orcompressibility of soil (e.g. triaxial compression tests, directshear tests, oedometer tests), and for measurement ofchemical properties of soil and groundwater (e.g. sulphate,

pH, etc).

Classification tests should always be carried out to provide general properties of the

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ground for pile design. Quantitative tests are necessary for assessing relevant designparameters if calculation methods based on soil and rock mechanics principles are used. Itmust be borne in mind that the design parameters obtained from laboratory testing relate tothose of the samples tested, and may therefore be subject to size effects, sample disturbance,sampling bias, etc.

As far as pile design is concerned, insitu tests can provide data for direct use in designby employing established semi-empirical correlations (e.g. results from SPT, CPT orpressuremeter tests). However, the applicability of such relationships to the particular fieldconditions must be carefully scrutinized. Alternatively, more fundamental soil or rockparameters, such as the angle of shearing resistance <£' , may be derived from the results ofinsitu tests, either through empirical correlations, e.g. relationship between SPT N value and<£' for sands (Peck et al, 1974), or directly from the interpreted test results by theory, e.g.pressuremeter (Mair & Wood, 1987).

Standard laboratory tests can provide data on design parameters, such as </> ' , for theassessment of skin friction and end bearing capacity of piles. Other special laboratory testssuch as direct shear tests to investigate the behaviour of interface between soil and steel orsoil and concrete may also be undertaken for pile design as appropriate (e.g. Johnston et al,1987; Lehane, 1992; Fahey et al, 1993). Oedometer testing is not commonly carried out onsaprolitic soils because of their fairly coarse-grained nature, particularly for granites, althoughoedometer testing may be more useful for clayey materials. In principle, stress path testingincorporating small strain measurements can be carried out to determine the yield loci andthe behaviour under different stress paths. Data from such high quality tests for soils inHong Kong are so far very limited because the tests are rarely required for routine piledesign.

2.8 ESTABLISHMENT OF GEOLOGICAL MODEL

An appropriate geological model of a piling site is essential to the pile design. Theinterpretation of borehole data, results of site mapping and available geological information,should be carried out by an experienced geotechnical engineer or engineering geologist toestablish a geological model that is suitable for engineering design.

There are inherent uncertainties in any geological models given that only a relativelysmall proportion of the ground can be investigated, sampled and tested. It is thereforeimportant that all available information is considered, such as nearby exposures andconstruction records of existing foundations, in characterizing the ground profile andcompiling a representative geological model for the site.

The representation on a borehole log of material, in a typical corestone-bearing rockmass weathering profile, uses the six-fold weathering grade classification for hand specimensnow generally followed in Hong Kong and based essentially on strength (GCO, 1988). Forgeneral engineering purposes, the geological model for a corestone-bearing jointed rock massshould be in terms of rock mass zones with differing proportions of relatively unweatheredmaterial, i.e. material grades I, II and III. Typical classification systems based on rock massgrades or classes are given in GCO (1988) and GCO (1990). However, it is customary inpractice to adopt a simple layered ground model, consisting of a planar rock surface overlain

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by successive soil layers, based on simplifications of the borehole logs in an attempt todelineate 'rockhead'. This concept needs to be viewed with extreme caution in a corestone-bearing profile as illustrated in Figure 2. The possibility of establishing an over-simplifiedgeological model, which may be incapable of reflecting the highly complex ground conditionsand potentially misleading, must be borne in mind. Continual vigilance during pileconstruction is called for, particularly in complex ground conditions such as weatheringprofiles and karst marble.

In view of the uncertainties and inherent variability of the weathering profile, thegeological model must be reviewed in the light of additional information. In this respect, theconstruction of each pile can be considered as a new stage of site investigation, to verify thegeological model.

The karst morphology in cavernous marble can be exceedingly complex. A detailedinvestigation is necessary in order to define a realistic geological model in Hong Kong thatis sufficient for design purposes. A classification system for cavernous marble rock masswhich facilitates the preparation of a geological model has been proposed by Chan et al(1994) as outlined in Section 5.11.

2.9 SELECTION OF DESIGN PARAMETERS

The selection of parameters for design of piles should take into account the quality andadequacy of the ground investigation, reliability of the geological model, the appropriatenessof the test methods in relation to the likely field conditions, the method of analysis adoptedfor the design, and the likely effects of pile construction on material properties. In principle,sophisticated analyses, where justified, should only be based on high quality test results. Thereliability of the output is, of course, critically dependent on the accuracy of the inputparameters.

'Best-estimate' parameters, which are those representative of the properties of thematerials in the field, should be selected for design. Guidance on the determination of 'best-estimate' parameters can be found in Geoguide 1 : Guide to Retaining Wall Design(GEO, 1993).

Engineering judgement is always required in the interpretation of test results and inthe choice of design parameters, having regard to previous experience and relevant casehistories. In adopting well-established correlations for a given geological material, it isimportant to understand how the parameters involved in the database for the particularcorrelation have been evaluated. In principle, the same procedure in determining theparameters should be followed to safeguard the validity of the correlations.

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3. TYPES OF PILE

3.1 CLASSIFICATION OF PILES

Piles can be classified according to the type of material forming the piles, the modeof load transfer, the degree of ground displacement during pile installation or the method ofinstallation.

Pile classification in accordance with material type (e.g. steel and concrete) hasdrawbacks because composite piles are available. A classification system based on the modeof load transfer will be difficult to set up because the proportion of skin friction and end-bearing that occurs in practice usually cannot be reliably predicted.

In the installation of piles, either displacement or replacement of the ground willpredominate. A classification system based on the degree of ground displacement during pileinstallation, such as that recommended in BS 8004 (BSI, 1986) encompasses all types of pilesand reflects the fundamental effect of pile construction on the ground which in turn will havea pronounced influence on pile performance. Such a classification system is thereforeconsidered to be the most appropriate.

In this document, piles are classified into the following four types :

(a) Large-displacement piles, which include all solid drivenpiles, including precast concrete piles, and steel or concretetubes closed at the lower end by a driving shoe or a plug,i.e. driven cast-in-place piles.

(b) Small-displacement piles, which include rolled steel sectionssuch as H-piles and open-ended tubular piles. However,these piles will effectively become large-displacement pilesif a soil plug forms.

(c) Replacement piles, which are formed by machine boring,grabbing or hand-digging. The excavation may need to besupported by bentonite slurry, or lined with a casing that iseither left in place or extracted during concreting for re-use.

(d) Special piles, which are particular pile types or variants ofexisting pile types introduced from time to time to improveefficiency or overcome problems related to special groundconditions.

This Chapter describes the types of piles commonly used in Hong Kong together withtheir advantages and disadvantages.

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3.2 LARGE-DISPLACEMENT PILES

3.2.1 General

The advantages and disadvantages of large-displacement piles are summarised inTable 2.

3.2.2 Precast Reinforced Concrete Piles

These piles are commonly in square sections ranging from about 250 mm to about450 mm with a maximum section length of up to about 20 m. Other pile sections mayinclude hexagonal, circular, triangular and H shapes, but these are not common in HongKong. Maximum allowable axial loads can be up to about 1 000 kN. The lengths of pilesections are often dictated by practical considerations including transportability, handlingproblems in sites of restricted area and facilities of the casting yard.

These piles can be lengthened by coupling together on site. Splicing methodscommonly adopted in Hong Kong include welding of steel end plates or the use of epoxymortar with dowels. Specially fabricated joints have been successfully used in othercountries, e.g. Scandinavia.

This type of pile is not suitable for driving into ground which contains a significantamount of boulders or corestones.

3.2.3 Prestressed Concrete Tubular Piles

Precast prestressed concrete piles commonly used in Hong Kong are closed-endedtubular sections of 400 mm to 600 mm diameter with maximum allowable axial loads up toabout 3 000 kN. Pile sections are normally 12 m long and are usually welded together usingsteel end plates. Pile sections up to 20 m can also be specially made.

Prestressed concrete piles require high-strength concrete and careful control duringmanufacture. Casting is usually carried out in a factory where the curing conditions can bestrictly regulated. Special manufacturing processes such as compaction by spinning orautoclave curing can be adopted to produce high strength concrete up to about 75 MPa.

A variant of the conventional precast prestressed concrete piles is described by Choy(1993). This involves a revised design of the pile shoe details aimed at improving thedriveability characteristics. Experience with this type of pile is limited.

Precast prestressed concrete piles may be handled more easily than precast reinforcedconcrete piles without damage. Spalling, cracking and breaking can occur if careful controlis not undertaken and good driving practice is not followed (see Section 7.2.5 for moredetails). This type of pile is generally less permeable than reinforced concrete piles and maybe expected to exhibit superior performance in a marine environment. Such piles have beensuccessfully employed in Hong Kong for many port works projects.

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3.2.4 Closed-ended Steel Tubular Piles

The use of box-section steel piles is not common in Hong Kong but steel tubular pilesare becoming increasingly popular, particularly for marine structures.

Steel tubular piles have high bending and buckling resistance, and have favourableenergy-absorbing characteristics for impact loading. Steel piles are generally not susceptibleto damage caused by tensile stresses during driving and can withstand hard driving. Drivingshoes can be provided to aid penetration.

The tubular piles may be infilled with concrete after driving, as appropriate.

3.2.5 Driven Cast-in-place Concrete Piles

Driven cast-in-place concrete piles are formed by driving a steel tube into the groundto the required set or depth and withdrawing the tube after concrete placement. The use ofprecast concrete rings instead of a steel tube is not common in Hong Kong. The tube maybe driven either at the top or at the bottom with a hammer acting on an internal concrete orcompacted gravel plug. A range of pile sizes is available, up to 600 mm in diameter. Themaximum allowable axial load is about 1 400 kN. The maximum length of such pilesconstructed in Hong Kong is about 30 m.

Proprietary systems of top-driven, cast-in-place piles have been used in Hong Kong.In this method, the steel tube is provided with a loose conical or flat cast-iron shoe whichkeeps the tube closed during driving. Light blows are usually imparted to the tube duringextraction, thus assisting concrete compaction.

For bottom-driven, cast-in-place piles with an expanded base, the tube does not haveto withstand direct impact and can be of a smaller thickness. Also, the piling rig does notneed to be as tall as rigs for other driven cast-in-place piling systems. When pile driving iscompleted, the tube is held against further penetration and the bottom plug is driven out bythe hammer within the tube. An enlarged pile base is formed using 'dry1 mix concrete, witha water/cement ratio of approximately 0.2, which is rammed heavily with the internalhammer.

3.3 SMALL-DISPLACEMENT PILES

3.3.1 General

Small-displacement piles are either solid (e.g. steel H piles) or hollow (open-endedtubular piles) with a relatively low cross-sectional area. This type of pile is usually installedby percussion methods; however a soil plug may be formed during driving, particularly withtubular piles, and periodic drilling out may be necessary to reduce the driving resistance. Asoil plug can create a greater driving resistance than a closed end, because of damping on theinner-side of the pile.

The advantages and disadvantages of small-displacement piles are summarised in

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Table 2.

3.3.2 Steel H-piles

Steel H-piles have been widely used in Hong Kong because of their ease of handlingand driving. Compared with concrete piles, they generally have better driveabilitycharacteristics and can generally be driven to greater depths. H-piles can be susceptible todeflection upon striking boulders, obstructions or an inclined rock surface. In areas underlainby marble, heavy-section H-piles with appropriate tip strengthening is commonly used topenetrate the karst surface and to withstand hard driving.

A range of pile sizes is available, with different grades of steel. The maximumallowable axial load is typically about 3 000 kN. Very large H sections (283 kg/m) with aworking load of about 3 600 kN have been used in some projects.

3.3.3 Open-ended Steel Tubular Piles

Driven open-ended tubular steel piles have been used in marine structures and inbuildings on reclaimed land. This type of pile has been driven to over 50 m. A plug willform when the internal skin friction exceeds the end-bearing resistance of the entire cross-sectional area the pile. Driving resistance can be reduced by pre-boring or by reaming outthe plug formed within the pile. Typical diameters range from 275 mm to about 2 m witha maximum allowable axial load of about 7 000 kN. Maximum pile diameter is oftengoverned by the capacity of the driving machine available, which is generally not greater thanDelonag 100 diesel hammer in Hong Kong.

3.4 REPLACEMENT PILES

3.4.1 General

Replacement, or bored, piles are formed by excavation or boring techniques. Whenconstructed in water-bearing soils which are not self-supporting, the pile bore will need to besupported using steel casing, concrete rings or drilling fluids such as bentonite slurry,polymer mud, etc. Excavation of the pile bore may also be carried out by hand-digging inthe dry; and the technique developed in Hong Kong involving manual excavation is knownlocally as hand-dug caissons.

The advantages and disadvantages of replacement piles are summarised in Table 3.

3.4.2 Machine-dug Piles

(1) General. Machine-dug piles are formed by rotary boring, or percussive methodsof boring, and subsequently filling the hole with concrete. Piles', with 600 .mm or less indiameter are commonly known as small-diameter piles. Piles greater than 600 mm diameterare referred to as large-diameter piles.

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(2) Small-diameter bored piles. This type of pile has sizes ranging from about300 mm to 600 mm with working loads up to about 1 500 kN.

One proprietary form of small-diameter bored pile involves the use of drop tools forexcavation and compressed air to compact the concrete in the pile shaft. The common sizesof this type of piles used in Hong Kong range from 325 mm to 508 mm, with working loadsup to about 1 000 kN. These piles can be installed in sites where the headroom is limited.These piles are sometimes constructed without reinforcement and the integrity of such un-reinforced piles when subject to ground movements arising from adjacent constructionactivities should be considered.

Another proprietary piling system is the continuous-flight auger (cfa) type piles suchas the 'Pakt-in-Place (PIP) Pile1 used in Hong Kong. In this system, the bore is formed usinga flight auger and concrete or grout is pumped in through the hollow stem. Sizes of PIP pilesrange from 300 mm to 700 mm diameter and lengths are generally less than 30 m. The cfapiles have considerable advantages over conventional bored piles in water-bearing andunstable soils by eliminating the need of casing and the problems of concreting underwater.The piles can be installed with little noise and vibration and are therefore suited for sites inurban areas. However, this type of pile cannot cope with boulders. The lack of penetrationunder continuous rotation due to a hard layer or an obstruction can lead to soil flighting upthe auger causing ground loss and settlement.

(3) Large-diameter bored piles. Large-diameter machine-dug piles are used in HongKong to support heavy column loads of tall buildings (e.g. Holt et al, 1982; Fraser, 1985)and highways structures such as viaducts (Fraser & Kwok, 1986). Typical sizes of these pilesrange from 1 m to 2.5 m, with lengths up to about 80 m and working loads up to about36 000 kN. Special mechanical tools are readily available for belling out the base. Pile oflarger diameters have been constructed in Hong Kong, but only to fairly shallow depth.

Traditionally in Hong Kong, large-diameter machine-dug piles are required to be'founded on rock1 with a prescribed safe bearing pressure. In reality, for many such boredpiles constructed in saprolite, the load is resisted primarily by skin friction.

3.4.3 Hand-dug Caissons

(1) General Hand-dug caissons have been used widely in Hong Kong as foundationsor earth retaining structures with a diameter typically ranging from 1.5 m to 2.5 m, and anallowable load of up to about 25 000 kN. Hand-dug caissons of a much larger size, ofbetween 7 m and 10 m in diameter, have also been constructed successfully (e.g. Humphesonet al, 1986; Barcham & Gillespie, 1988). The advantages and disadvantages of hand-dugcaissons are summarised in Table 3.

Hand-dug caisson shafts are excavated using hand tools in stages with depths of upto about 1 m, depending on the competence of the ground. Dewatering is facilitated bypumping from sumps on the excavation floor or from deep wells. Advance grouting may becarried out to provide support in potentially unstable ground. Each stage of excavation islined with insitu concrete rings (minimum 75 mm thick) using tapered steel forms whichprovide a key to the previously constructed rings. When the diameter is large; the rings may

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be suitably reinforced against stresses arising from eccentricity and non-uniformity in hoopcompression. Near the bottom of the pile, the shaft may be belled out to enhance thetheoretical load-carrying capacity.

The isolation of the upper part of hand-dug caissons by sleeving is sometimesprovided for structures built on sloping ground to prevent the transmission of lateral loads tothe slope or conversely the build-up of lateral loads on caissons by slope movement (GCO,1984). However, there is a lack of instrumented data on the long-term performance of thesleeving.

(2) Technical guidance. Given the disturbingly high accident rate and the healthhazard in installation and construction (Ng et al, 1987) of caissons, their use should bediscouraged. Examples of situations where the use of caissons should be avoided include :

(a) coastal reclamation sites with high ground water table,

(b) sites underlain by cavernous marble,

(c) deep foundation works (e.g. in excess of say 50 m),

(d) landfill or chemically-contaminated sites,

(e) sites with a history of deep-seated ground movement,

(f) sites in close proximity to water or sewerage tunnels,

(g) sites in close proximity to shallow foundations, and

(h) sites with loose fill having depths in excess of say 10 m.

Hand-dug caissons should be constructed only when they are the only practicablesolution and there is no safe engineered alternative. All reasonable and appropriateprecautionary measures must be taken to ensure adequate protection of workers.

Examples of situations where hand-dug caissons may be considered include :

(a) steeply-sloping sites with hand-dug caissons of less than25 m in depth in soil, and

(b) sites with difficult access or insufficient working roomwhere it may be impracticable or unsafe to use mechanicalplant.

In all cases, the desirable minimum internal diameter of hand-dug caissons is 1.8 m.

Before opting for hand-dug caissons, a risk assessment should be carried out coveringgeneral safety, the cost of damage arising from de water ing, and the possibility of unforeseenground conditions. The design of caisson linings should also be examined for suitability asfor any other structural temporary works.

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A guide to good practice for the design and construction of hand-dug caissons has beenproduced by the Hong Kong Institution of Engineers (1981). Further discussion on thepotential problems during construction of hand-dug caissons is given in Section 7.4.3.

Where hand-dug caissons are employed, consideration should be given to the followingprecautionary measures and preventive works, as appropriate :

(a) carrying out additional ground investigation to obtain bestpossible information about the ground conditions,

(b) pre-grouting around each hand-dug caisson to reduce therisk of collapse and limit the groundwater drawdown,

(c) installation of cut-off walls or curtain grouting around thesite boundary or around groups of caissons to limit inflowof water,

(d) installation of dewatering wells within the site, possiblysupplemented by recharge wells around the periphery ofthe site to limit the groundwater drawdown in adjacentground,

(e) construction of the caissons in a suitable sequence,

(f) reduction in the depth of each caisson digging stage,

(g) provision of immediate temporary support for theexcavated face prior to the casting of the concrete liner,

(h) provision of steel reinforcement to the concrete liner,

(i) driving dowels radially into the surrounding soil asreinforcement at the bottom of excavation to reduce thechance of heaving,

(j) provision of a drainage or relief well at the position ofeach caisson in advance of manual excavation,

(k) avoidance of the introduction of new caisson gangs intopartly-completed excavations,

(1) completion of proper grouting of ground investigationboreholes and old wells in the vicinity of hand-dugcaissons,

(m) provision of good ventilation,

(n) use of well-maintained and checked equipment,

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(o) safety inspections,

(p) provision of safety equipment,

(q) an assessment of the risks by a safety professional to thehealth and safety of the workers whilst at work in caissonsand implementing, monitoring and reviewing the measuresto comply with the requirements under all existing safetylegislation,

(r) monitoring and control of the potential health hazards,e.g. poisonous gases, oxygen deficiency, radon and silicadust, and

(s) monitoring the ground water table and possibly the groundand sub-soil movement by piezometers and inclinometersinstalled around the site boundary.

For general guidance on the practicable safety and health measures in the constructionof hand-dug caissons, reference may be made to the 'Code of Safe Working Practices forHand-dug Caissons' published by the Occupational Safety & Health Council (1993).

(3) Engagement of suitable staff. One of the most important elements in the successof a hand-dug caisson project is the engagement of suitably qualified and experiencedprofessionals in the geotechnical assessment and investigation of the site to identify potentiallyunfavourable geological and hydro-geological conditions that may give rise to engineering andconstruction problems, and to implement the necessary precautionary and preventivemeasures. Likewise, the employment of suitably trained and experienced constructionworkers, together with adequate supervision to promote strict adherence to stringent safetyand health requirements, is also a pre-requisite.

3.5 SPECIAL PILE TYPES

3.5.1 General

Three special pile types, viz. barrettes, mini-piles and composite piles, are discussedbelow.

3.5.2 Barrettes

A barrette of rectangular section is a variant of the traditional bored pile. Therectangular holes are excavated with the use of grabs. In Hong Kong, common barrette sizesare 0.8 m x 2.2 m and 1.2 m x 2.8 m, although the panels can be up to about 5 m to 6 m,with depths to about 80 m. Because of their rectangular shape, barrettes can be oriented togive maximum resistance to moments and horizontal forces.

Load tests on barrettes founded in saprolite have demonstrated that significant skin

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friction can also be mobilised (e.g. Pratt & Sims, 1990).

3.5.3 Mini-piles

Mini-piles generally have a diameter of between 100 mm and 250 mm. Reinforcementbars are provided in the piles.

Construction can be carried out typically to about 60 m depth or more, althoughvertically control will become more difficult at greater depths. Mini-piles are usually formedby small drilling rigs with the use of down-the-hole hammers or rotary percussive drills.They can be used for sites with difficult access or limited headroom and for underpinning.In general, they can overcome large or numerous obstructions in the ground.

Given the small diameter and high slenderness ratio of mini-piles, the load is resistedlargely by skin friction. Where mini-piles are installed in soil, the working load is usuallyless than 700 kN but can be in excess of 1 000 kN if post grouting is undertaken using tube-a-manchette. In the case of mini-piles forming rock sockets, the capacity may be limited bythe prescribed permissible structural stresses, which may be up to about 1 350 kN.

The structural design of mini-piles is discussed in Sections 5.12.4 & 5.12.5. Wheremini-piles are employed, the combined resistance of the pile cap and the piles may be usedto take the horizontal loading. Comments on this design approach are given inSection 6.5.2(3).

Corrosion aspects of mini-piles are discussed in Section 5.14.

3.5.4 Composite Piles

Some systems of composite piles have been developed to deal with special siteconditions. Three types of composite piles that have been used in Hong Kong are discussedbelow.

The first type is essentially a combination of driven cast-in-place techniques with pre-formed pile sections in reclamations. In this system, a driven cast-in-place piling tube isinstalled and the expanded base is concreted. A steel H-section is then inserted and beddedusing light hammer blows. Further concrete is introduced to provide a bond length sufficientto transfer the load from the steel section. The concrete is terminated below the soft depositsand the remainder of the piling tube is filled with sand before it is extracted.

Similar composite construction has also been tried with other driven cast-in-place pilingsystems in combination with precast concrete sections, which may be sleeved with bitumen,in order to avoid the risk of damage to the coating during driving. A variant of this type ofcomposite pile involves grouting steel H-sections into sockets drilled in rock, instead of usingrock anchors or dowels, to resist uplift forces (Construction & Contract News, 1993).

The second type of composite pile is the Steel-Concrete Composite (SC) Pile. Thiscomprises a structural steel casing with a hollow spun concrete core and a solid driving shoe.

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By combining the advantages of good quality concrete and high strength external steel pipecasing, SC pipe piles can provide better driveability and lateral load resistance but moreemphasis has to be placed on corrosion protection. Pile sizes are similar to precastprestressed piles with maximum working loads of about 2 800 kN. The piles can be installedwith the centre-augering system (Fan, 1990) which is a non-percussive system with minimalnoise and vibrations. The augering and drilling can be carried out in the centre hole of thepile which is jacked into the predrilled hole by a counter weight and hydraulic jack mountedon the machine. The final set can be obtained using a pile driving hammer.

A third type of composite pile is the drill-and-drive system whereby the open-endedtubular pile is drilled out when it cannot be driven further. Such a system may, in principle,be used to facilitate penetration of cavernous marble in Hong Kong. However, it is importantto exercise stringent control on the drilling procedure to avoid excessive loss of ground.

If concrete is cast into a steel tube after it has been driven, the allowable capacity ofthe composite pile will be influenced by strain compatibility requirements. Considerationshould be given to the possible effect of radial shrinkage of the concrete which can affect thebond with the steel tube. Shear keys may be used to ensure adequate shear transfer in thecase where the upper part of an open-ended steel tube is concreted (Troughton, 1992).

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4. CHOICE OF PILE TYPE AND DESIGN RESPONSIBILITY

4.1 GENERAL

This Chapter provides guidance on the factors that should be considered in choosingthe most appropriate pile type. Issues relating to the allocation of design responsibility arealso discussed.

4.2 FACTORS TO BE CONSIDERED IN CHOICE OF PILE TYPE

4.2.1 General

The determination of the need to use piles and the identification of the range offeasible pile types for a project form part of the design process. In choosing the mostappropriate pile type, the factors to be considered include ground conditions, nature ofloading, effects on surrounding structures and environs, site constraints, plant availability,safety, cost and programme, taking into account the design life of the piles.

Normally, more than one pile type will be technically feasible for a given project andthe selection process is in essence a balancing exercise between various, and sometimesconflicting, requirements. The choice of the most suitable type of pile is usually reached byfirst eliminating any technically unsuitable pile types followed by careful consideration of theadvantages and disadvantages of the feasible options identified. Due regard has to be paidto technical, economical, operational, environmental and safety aspects. A flow chartshowing the various factors to be considered in the selection of piles is given in Figure 3.

It should be noted that possible installation problems associated with the different piletypes should not be the sole reason for rejection as these can generally be overcome byadherence to good piling practice and adoption of precautionary measures, albeit at a cost.However, from a technical viewpoint, the choice of piles should be such as to minimisepotential construction problems in the given site and ground conditions, and limit the risk ofpossible delays. Delays are especially undesirable where the project owner is payingfinancing cost.

4.2.2 Ground Conditions

(1) General The choice of pile type is, in most instances, affected by the prevailingground conditions. The presence of obstructions, soft ground, depth of founding stratum,cavities, faults, dykes, aggressive ground, etc. will have a significant influence on thesuitability of each pile type.

(2) Obstructions. Problems caused by obstructions are common in old reclamations,public dump sites, and ground with bouldery colluvium or corestones in saprolites. Drivenpiles are at risk of being deflected or damaged during driving. Additionally, it is possiblethat penetration of large-displacement piles may be limited in the case where groundimprovement works have been carried out in fills such as hydraulically-placed sands.

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Measures that can be adopted to overcome obstructions are described in Sections 7.2.5(4) and7.3.4(4).

(3) Soft ground. In soft ground, such as marine mud or organic soils, cast-in-placepiles can suffer necking unless care is taken when extracting the temporary casing.Construction of hand-dug caissons can be particularly hazardous because of possible pipingor heaving at the base. Machine bored piles with permanent casings can be used to alleviateproblems of squeezing.

In these ground conditions, driven piles offer benefits as their performance is relativelyindependent of the presence of soft ground.

(4) Depth of founding stratum. The depth of the founding stratum can dictate thefeasibility of certain pile types. Advance estimates of the depth at which a driven pile islikely to reach a satisfactory 'set' are usually made from a rule-of-thumb which relies on SPTresults. The SPT N value at which concrete piles are expected to reach 'set' is quoted bydifferent writers in Hong Kong in the range of 50 to 100, whilst the corresponding N valuefor steel H-piles to reach set is quoted as two to three times greater.

Barrettes and large-diameter machine-dug piles are generally limited to depths of 60 mto 80 m although equipment capable of drilling to depths in excess of 90 m is readilyavailable.

(5) Cavernous marble. Parts of the Yuen Long and Ma On Shan areas are underlainby marble formations. The upper surface of marble can be karstic and deep cavities may alsobe present. The assessment of piling options requires a careful consideration of the karstmorphology.

Experience in Hong Kong indicates that heavy section steel H-piles (e.g. 305 mm x305 mm x 186 kg/m or 223 kg/m) with reinforced tips can generally penetrate the surfacekarst zone under hard driving. However, pre-drilling may have to be adopted for sites withunfavourable karst features such as large overhangs. Large-diameter bored piles have alsobeen constructed through cavernous marble (Li, 1992).

Precast concrete piles are prone to being deflected where the rock surface is steeplyinclined or highly irregular and may suffer damage under hard driving. Most types of drivencast-in-place piles are unsuitable because of difficulty in seating the piles in sound marble.

The use of hand-dug caissons should be avoided because of the risk of sinkholesinduced by dewatering and potential inrush of soft cavity infill. Barrettes may be difficultto construct because of the possibility of sudden loss of bentonite slurry through opencavities.

(6) Faults, shear zones or dykes. In sites traversed by faults, shear zones or dykes,the geology and the weathering profile can be highly variable and complex. Dykes areespecially common in the Sung Kong Granite, Ma Wan Granite and Sha Tin GraniteFormations in the western part of Hong Kong (Strange & Shaw, 1986), The choice of pileswill be affected by the need to cope with variable ground conditions and the feasibility of thedifferent pile types will be dependent on the capability of the drilling equipment or

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driveability considerations.

(7) Aggressive or contaminated ground. Corrosion of piles should be a particulardesign consideration in situations such as those involving acidic soils, industrial contaminants,the splash zone of marine structures and in ground where there is a fluctuating groundwaterlevel (Section 5.14). In general, precast concrete piles, which allow stringent quality controland the use of high strength material, are preferred in aggressive or contaminated ground.

4.2.3 Nature of Loading

Pile selection should take into account the nature and magnitude of the imposed loads.In circumstances where individual pile spacing could result in the problem of 'pile saturation',the use of large-diameter replacement piles may need to be considered.

For structures subject to cyclic and/or impact lateral loading such as in jetties and quaystructures, driven steel piles may be suitable as they have good energy-absorbingcharacteristics.

In the case of large theoretical lateral loads (e.g. tall buildings), piles with a highmoment of resistance may have to be adopted.

4.2.4 Effects of Construction on Surrounding Structures and Environment

(1) General. The construction of piles can cause damaging or disturbing effects onsurrounding structures and environs. These should be minimised by the use of appropriatepile type and construction methods. The constraints that such effects may pose on the choiceof pile type vary from site to site, depending on ground conditions and the nature ofsurrounding structures and utilities.

(2) Vibration. Vibrations caused by piling are a nuisance to nearby residents andcould cause damage to utilities, electronic equipment and vulnerable structures such asmasonry works. Large-displacement piles are likely to produce greater ground vibration thansmall-displacement and replacement piles.

(3) Noise. Construction activities, including percussive piling, are subject to theprovisions of the Noise Control Ordinance (Hong Kong Government, 1994). Percussivepiling is banned within the restricted hours, i.e. from 7 p.m. to 7 a.m. on weekdays andwhole day on Sundays and public holidays, and restricted to three or five hours duringweekdays in densely-populated areas. It should be noted that the Government intends tophase-out diesel hammers, which are very noisy and prone to emit dark smoke, forenvironmental reasons.

(4) Groundwater drawdown. Excavation of hand-dug caissons below the groundwatertable requires dewatering. The resulting ground movements may seriously affect adjacentutilities, roads and structures supported on shallow foundations.

(5). '-Rise'' in groundwater. Closely-spaced piles below the groundwater may dam

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groundwater flow leading to a rise in groundwater levels (Pope & Ho, 1982). This may beparticularly relevant for developments on steeply-sloping hillsides, especially where groutinghas been carried out, e.g. in hand-dug caisson construction. The effect of rise in groundwateron adjacent underground structures like MTR tunnels, e.g. increase in buoyancy, should alsobe considered.

(6) Ground heave. Installation of displacement piles will result in heave and lateraldisplacement of the ground, particularly in compact fine-grained sandy silts and clayey soils(Malone, 1990), and may affect adjacent structures or piles already installed. The use ofreplacement piles will obviate such effects. Should displacement piles be used for otherreasons, precast piles, as opposed to driven cast-in-place piles, may be considered as theyoffer the option that uplifted piles can be re-driven.

• (7) Pollution. Spoil and contaminated drilling fluid, for replacement pileconstruction, especially those arising from reclamation area, cause nuisance to surroundingenvironment and would need to be properly disposed of (Environmental ProtectionDepartment, 1994).

4.2.5 Site and Plant Constraints

In selecting pile types, due consideration should be given to the constraints posed bythe operation of the equipment and site access.

All driven piles and machine-dug piles require the use of large rigs. These mayrequire substantial temporary works for sloping ground and sites with difficult access.

Headroom may be restricted by legislation (e.g. sites near airports) or physicalobstructions such as overhead services. In such case, large crane-mounted equipment maynot be appropriate and mini-piles will be a feasible option.

The construction of replacement piles may involve the use of drilling fluid. Theancillary plant may require considerable working space. On the other hand, prefabricatedpiles similarly will require space for storage and stockpiling. These two types of piles maytherefore cause operational problems on relatively small sites.

4.2.6 Safety

Safety considerations form an integral part in the assessment of method ofconstruction. Problems with hand-dug caissons include inhalation of poisonous gas and silicadust by workers, insufficient ventilation, base heave, piping, failure of concrete linings andfalling objects (Chan, 1987), and their use is strongly discouraged in general.

Installation of machine-dug piles and driven piles requires heavy piling rigs. Accidentsinvolving collapse or overturning of the rigs, that can be caused by overloading, swingingloads, incorrect operation, wind gusts or working on soft or steeply-sloping ground, can resultin casualties. Serious accidents may also occur when loads swing over personnel as a resultof failure of chain or rope slings due to overloading, corrosion or excessive wear.

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Notwithstanding the safety risks and hazards involved in pile construction, it shouldbe noted that most of these can be minimised provided they are fully recognized at the designstage and reasonable precautions are taken and adequate supervision provided. Vetting ofcontractor's method statements provides an opportunity for safety measures to be included inthe contract at an early stage.

4.2.7 Programme and Cost

The design engineer frequently has a choice between a number of technically feasiblepiling options for a given site. The overall cost of the respective options will be a significantconsideration.

The scale of the works is a pertinent factor in that high mobilisation costs of largeequipment may not be cost effective for small-scale jobs. The availability of plant can alsoaffect the cost of the works. Contractors may opt for a certain piling method, which may notbe the most appropriate from a technical point of view, in order to optimise the material,equipment and plant available to them amongst the ongoing projects.

The cost of piling in itself constitutes only part of the total cost of foundation works.For instance, the cost of a large cap for a group of piles may sometimes offset the higher costof a single large-diameter pile capable of carrying the same load. It is necessary to considerthe cost of the associated works in order to compare feasible piling options on an equal basis.

A most serious financial risk in many piling projects is that of delay to projectcompletion and consequential increase in financing charges combined with revenue slippage.Such costs can be much greater than the value of the piling contract. The relativevulnerability to delay due to ground conditions, therefore, ought to be a factor in the choiceof pile type.

4.3 DESIGN RESPONSIBILITY

4.3.1 Contractor's Design

Traditionally in Hong Kong, "Contractor's design" is the favoured contractual optionfor piling works. Under this system, the professional engaged by the client as the projectdesigner provides the tenderers with the relevant information. This includes information onground conditions, loading, acceptance criteria of the piles in the required load tests, togetherwith specific constraints on noise, vibration, headroom, access, pile length and vertically,etc, as appropriate. The project designer may, in some instances, choose to rule out thosepile types that are obviously unsuitable for the project in the specification.

Under this arrangement, the contractor is required to choose the pile type and designthe layout of the piles (sometimes including the pile caps). The contract is usually based ona lump sum under which the contractor undertakes to install the piles to meet the acceptancecriteria and is required to bear all the risks in respect of design, construction, cost andprogramme of the works.

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4.3.2 Engineer's Design

Under "Engineer's design", the design responsibility rests with the project designer.This is the common approach for piling works in Government civil engineering contracts andlarge private building developments. The methods of construction will not be specified indetail but good construction practice and quality control requirements are usually included inthe specifications. The project designer will also supervise pile construction and monitorquality control tests, check the general compliance of the works with the specification and thedrawings, assess the adequacy of the founding depth of each pile, and verify his designassumptions against field observations.

Where the piles are designed by the project designer, the assumptions made in thedesign, together with the ground investigation information, should be communicated to thetenderers. The method of construction selected by the contractor must be compatible withthe design assumptions. It is essential that the designer is closely involved with the siteworks to ensure that the agreed construction method is followed and that the necessary designamendments are made promptly.

The contractor is responsible for the workmanship and method of construction, andis required to provide adequate supervision to ensure adherence to the agreed methodstatement. Under this arrangement, the re-measurement form of contract is generally adoptedand the contractor is reimbursed agreed costs arising from variations.

The tenderers for a piling contract are usually allowed to submit alternative designsin order that a more cost-effective or suitable solution will not be overlooked. The alternativedesign will be subject to the agreement of the project designer. In practice, it is usual toundertake preliminary enquiries with potential specialist piling contractors prior to tenderingto discuss the range of suitable piling options given the specific constraints on the project.This is particularly useful if the range of specialist piling contractors can be nominated by theproject designer, and can help to avoid the submission of technically unsuitable alternativeproposals.

4.3.3 Discussion

The benefits of the approach based on "Contractor's design" include the following

(a) The contractor's experience, technical expertise and hisknowledge on availability and costs of material, plant andlabour associated with a particular pile type can be utilised.Aspects of buildability can be properly assessed by thecontractor, particularly where proprietary piling systems areinvolved.

(b) There is comparatively less ambiguity in terms of therespective liability of the project designer and the contractorfor the performance of the works.

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The benefits of the approach based on "Engineer's design" include the following

(a) Engineers, when choosing the pile type, may be moreobjective and are less likely to be restricted by plantavailability and past expertise in certain pile types, andtherefore the best overall piling system will be considered.

(b) Engineers are less influenced by cost considerations and canconcentrate more on the technical grounds. For projects indifficult site and ground conditions requiring significantengineering input, the use of the "Engineer's design"approach is particularly warranted. This is because thecontractor's chosen scheme may involve undue risk offailing to comply with the specified performance criteria.

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5. DESIGN OF SINGLE PILES AND DEFORMATION OF PILES

5.1 GENERAL

In Hong Kong, permissible soil and material stresses are prescribed in Regulations andCodes for the design of piles. In traditional local building practice, the settlement of the pilefoundation is customarily not checked, with the implicit assumption that the settlement of abuilding with piles provided in accordance with the design rules will be tolerable. Empiricalpile design rule works well within the database on which it has been developed. When newdesign requires extrapolating past experience beyond the database, such empirical design maybe either needlessly expensive or unconservative.

Methods based on engineering principles of varying degrees of sophistication areavailable as a framework for pile design. All design procedures can be broadly divided intofour categories :

(a) empirical frules-of-thumb',

(b) semi-empirical correlations with insitu test results,

(c) rational methods based on simplified soil mechanics or rockmechanics theories, and

(d) advanced analytical (or numerical) techniques.

A judgement has to be made on the choice of an appropriate design method for a givenproject. In principle, in choosing an appropriate design approach, relevant factors that shouldbe considered include :

(a) the ground conditions,

(b) nature of the project, and

(c) comparable past experience.

This Chapter covers the design philosophies including recommended factors of safetyand outlines the various design methods for single piles. Emphasis is placed on pile designmethods in granular soils given that granitic soils are generally regarded as granular soils incurrent Hong Kong practice as far as their general engineering behaviour is concerned.Appropriate design methods for piles in rock, karstic conditions and clays are also outlined.Recommendations are given on the appropriate pile design methods that may be adopted foruse in Hong Kong.

5.2 PILE DESIGN IN RELATION TO GEOLOGY

Geological input is crucial in foundation works and should commence at an early stageof planning of a project. The geology of Hong Kong has been briefly described in

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Section 2.2.3. The importance of a representative geological model in the design of pilefoundations is highlighted in Section 2.8.

Theoretical methods of pile design have been developed for simple cases such as pilesin granular soil, or piles on rock. Judgement should be exercised in applying the simplifiedpile design methods, having regard to past experience with the use of these methods inspecific local geological conditions.

5.3 DESIGN PHILOSOPHIES

5.3.1 General

The design of piles should comply with the following requirements throughout theirservice life :

(a) There should be adequate safety against ultimate limit statefailure of the ground. The required factor of safety dependson the importance of the structure, consequence of failure,reliability and adequacy of information on groundconditions, sensitivity of the structure, nature of the loading,local experience, design methodologies, number ofrepresentative preliminary pile load tests, etc.

(b) There should be adequate margin against excessive pilemovements which would impair the serviceability of thestructure.

5.3.2 Global Factor of Safety Approach

The conventional global factor of safety approach is based on the use of a lumpedfactor applied notionally to either the ultimate strength or the applied load. This is deemedto cater for all the uncertainties inherent in the design.

The conventional approach of applying a global safety factor provides for variationsin loads and material strengths from their estimated values, inaccuracies in behaviouralpredictions, unforeseen changes to the structure from that analyzed, unrecognised loads andground conditions, errors in design and construction, and acceptable deformations in service.

5.3.3 Partial Factor of Safety Approach

In the past two decades, design of concrete structures has largely been based on partialfactors, i.e. different safety factors are defined for each type of material and loading to reflectthe relative uncertainties.

In the partial factor of safety approach for the design, of foundations, the design loadand design strength are obtained by applying the respective partial safety factors. The

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governing criterion is that the design loading must not induce stresses that exceed the designstrength.

In principle, partial factors may be applied to the strength parameters or the pileresistance calculated based on unfactored strength parameters.

5.3.4 Discussion

If it is accepted that pile design is often governed by movements, it will be somewhatacademic to debate the pros and cons of global versus partial factor methods of design, andwhat may be needed is a movement-based design. However, this would require extensiveknowledge of the range of load-deformation behaviour of different types of piles. It iscustomary to check for a sufficient margin of safety before undertaking a serviceability check.

In general, the partial safety factor approach has advantages over the conventionalapproach in terms of distribution of safety margins, explicit consideration of uncertainty andcompatibility with other codes (Simpson, 1992). Potential therefore exists in the partial factorapproach for determining the margin of safety in a more logical manner.

In the case of piling, there is the fundamental need to consider strain compatibility asa result of the difference in the rate of mobilisation of shaft friction and end bearingresistance. Much larger movements are required to fully mobilise the end bearing capacitythan the skin friction. Thus, under working load, the proportion of mobilised skin frictionand base resistance will be different, amounting effectively to different partial factors. Thesefactors, which are governed by the limiting movement at working load conditions, may betaken to be 'serviceability' or 'mobilisation' factors.

However, pile head movements will be a function of the loading and pile-soil stiffness,which could vary by one or two orders of magnitude and can be significantly affected by thepile construction method and workmanship. An added complication is that there are amultitude of design methods, many of which are semi-empirical in nature. Furthermore, thebehaviour of piles is affected by soil-structure interaction, and it is difficult to achievecompatibility between the partial factors for geotechnical design and those for structuraldesign. There is ongoing debate on the appropriate load factors to be used for geotechnicaldesign (Simpson, 1993).

In theory, the partial factor approach is more rational than the global factor approach.It is possible to calibrate the partial factors against the global safety factors based onsimplified reliability index calculations. The framework for partial factors, given theuncertainties in the loading, resistance, method of analysis, construction technique, etc, canhowever be exceedingly complex for piling. In view of a lack of adequate data onperformance of piles in Hong Kong conditions, such a system will offer no real advantageover the existing system in practice.

In summary, it is suggested that the conventional use of global safety factors, temperedwith suitable designer's judgement, should be adequate in practice for the time being.

For practical purposes, it is recommended that piles are designed on the basis of an

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adequate global factor of safety against ultimate failure. An additional check should be madeusing minimum 'mobilisation' factors to ensure there is a sufficient margin against excessivemovement of the pile. It is necessary to estimate the deformation of the foundation toconfirm that the serviceability requirements including total and differential movements aremet.

5.3.5 Recommended Factors of Safety

The following considerations should be taken into account in the selection of theappropriate factors of safety :

(a) There should be an adequate safety factor against failure ofstructural members in accordance with appropriate structuralcodes.

(b) There must be an adequate global safety factor on ultimatebearing capacity of the ground. Terzaghi and Peck (1967)proposed the minimum acceptable factor of safety to bebetween 2 and 3 for compression loading. The factor ofsafety should be selected with regard to importance ofstructure, consequence of failure, the nature and variabilityof the ground, reliability of the calculation method anddesign parameters, extent of previous experience andnumber of load tests on preliminary piles. The factors assummarised in Table 4 for piles in soils should be appliedto the sum of the shaft and base resistance.

(c) The assessment of working load should additionally bechecked for minimum 'mobilisation' factors fs and f^ on theskin friction and base resistance respectively as given inTable 5.

(d) • Settlement considerations, particularly for sensitivestructures, may govern the allowable loads on piles and theglobal safety factor and/or 'mobilisation' factors may needto be higher than those given in (b) & (c) above.

(e) Where significant cyclic, vibratory or impact loads areenvisaged or the properties of the ground are expected todeteriorate significantly with time, the minimum globalfactor of safety to be adopted may need to be higher thanthose in (b), (c) and (d) above.

The minimum factors of safety recommended for pile design are intended to be usedin conjunction with best estimates of resistance (Section 2.9).

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5.4 AXIALLY LOADED PILES IN SOIL

5.4.1 General

In the evaluation of the ultimate bearing capacity of an axially loaded pile in soil (incorestone-bearing weathering profiles, 'soil' may be taken as zones with a rock content notmore than 50%), a number of methods are available :

(a) pile driving formulae for driven piles,

(b) wave equation analysis for driven piles,

(c) calculation methods based on simplifying soil and rockmechanics principles,

(d) correlation with Standard Penetration Tests (SPT), and

(e) correlation with other insitu tests such as cone penetrationtests and pressuremeter tests.

The satisfactory performance of a pile is, in most cases, governed by the limitingacceptable deformation under various loading conditions. Hence, the settlement of pilesshould be checked where appropriate. Reference may be made to Section 5.13 for therecommended methods of assessing movements.

In addition to the above methods, the design of piles can also be based on results ofload tests. This is discussed in Section 5.10.

5.4.2 Pile Driving Formulae

Pile driving formulae relate the ultimate bearing capacity of driven piles to the finalset (i.e. penetration per blow). Various formulae have been proposed, such as the HileyFormula or Dutch Formula, which are generally based on the principle of conservation ofenergy. The inherent assumptions made in some formulae pay little regard to the actualforces which develop during driving, or the nature of the ground and its behaviour.

Chellis (1961) observed that some of these formulae were based on the assumptionsthat the stress wave due to pile driving travels very fast down the pile and the associatedstrains in the pile are considerably less than those in the soil. As a result the action of theblow is to create an impulse in the pile which then proceeds to travel into the ground as arigid body. Where these conditions are fulfilled, dynamic formulae give good predictions.As noted by Chellis,-if'the' set becomes small such that the second condition is not met, thenthe formulae may become unreliable.

In Hong Kong, the Hiley Formula has been widely used for the design of driven piles.The formula is as follow :

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R = .................. (5.1)d ( s * + c * / 2 )

where Rd = driving resistance7jh = efficiency of hammer (allowing for energy loss on impact)

Wh = weight of hammerdh = height of fall of hammers * = permanent set of pilec * = elastic or recoverable movement of pile

A common requirement is that the final set should not be less than 2.5 mm/blow,unless the pile is founded on rock, in an attempt to avoid over-driving. However, theconventional use of the Hiley formula implies that the only way to achieve this is to increasethe hammer energy which in turn may result in pile damage (Gammon, 1982).

A further problem is that by use of heavy hammers and hard driving, the theoreticalcapacity based on Hiley Formula may be significantly greater than that allowed using rationalmethods, particularly where set is achieved on either rock or by shallow embedment infounding stratum.

According to dynamic stress wave theory, it is not rational to take into account the fullweight of a pile in the Hiley Formula where the pile length exceeds about 30 m, dependingon the velocity of wave propagation through the pile. For very long piles, Cornfield (1961)proposed a modification of the Hiley Formula that involves assuming a constant effective pilelength instead of the full pile length. For such piles, it would be more rational, in principle,to undertake a wave equation analysis as described in Section 5.4.3 below.

Like other driving formulae, Hiley 's formula does not account for the rate at whichthe soil is sheared during pile driving and therefore cannot differentiate between cohesive andgranular soils. The high strain rates in cohesive soils during pile penetration can cause theviscous resistance of the soil to be considerably greater than the static capacity of the pile.Poskitt (1991) shows that without considering soil damping, the driving resistance can beoverestimated by several folds.

Problems may also occur where a pile is driven to a set on a corestone, overlyingmedium dense saprolite, or where depth of soil is thin so pile is driven to set on rock atshallow depth.

Despite various shortcomings of pile driving formulae, it appears that the HileyFormula has generally resulted in satisfactory foundations in Hong Kong as an adequate depthis usually achieved in fairly uniform soil profiles (Davies & Chan, 1981). However, this isnot the case for piles driven through thick layers of soft marine clays to the underlyingdecomposed rocks, and there are a number of cases in Hong Kong of large buildingsettlement and tilting occurring as a direct result of inadequate penetration of the piles intothe bearing stratum (Lumb, 1972; Lumb, 1979), Yiu & Lam (1990) noted from five pilesload-tested to failure that the comparison of the measured pile capacity with that predicted byHiley Formula was variable and inconsistent. Extreme caution should be exercised in placingtotal reliance on the use of pile driving formulae without due regard to the ground conditions.

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Given the potential shortcomings of pile driving formulae, it is recommended that thedesign of a driven pile should be supplemented by a 'static' capacity calculation, using eithersoil mechanics principles or established correlations with insitu tests (Sections 5.4.4 & 5.4.5).The design pile length should be taken as the greater of those calculated from a pile drivingformula and that from at least one 'static' method. Thus, if the Hiley set cannot be achievedwhen the pile is driven to the minimum design length calculated from 'static' method, the pilewill be required to be driven further until the Hiley set is achieved. Conversely, if therequired Hiley set is achieved at a higher elevation, an attempt will be made to drive the pilefurther, where possible without damaging the pile, until the minimum required penetrationhas been achieved. Where it is not possible to drive the pile further, the pile capacity shouldbe taken as the smaller of that calculated from the pile driving formula and that from at leastone 'static' method. If a driven pile is designed on the basis of a 'static' calculation withoutbeing driven to set, it is advisable that the design be confirmed by static load test.

Driving piles in dense soils may cause dilation which can result in false sets. This isparticularly liable to happen in dense silty sands or sandy silts when dilation causes negativepore water pressures followed by a reduction in end-bearing resistance when they aredissipated (see also Section 7.2.5(11)).

The final set of a pile, particularly where the piling formula has been calibrated againstsatisfactory static load test results and corresponding borehole information, will be useful asa site control measure. Experience suggests that driving to a target set pre-determined by apile driving formula can help to ensure no 'slack' in the pile-soil system compared to the caseof driving the pile to a pre-determined length only.

Test driving may be considered at the start of a driven piling contract to assess theexpected driving characteristics.

5.4.3 Wave Equation Analys is

A wave equation analysis based on the theory of wave propagation (Figure 4 andSection 8.4.3) can be undertaken to assess pile behaviour during driving. The reliability ofthe results depends on the appropriateness of the model and the accuracy of the input data,including the ground properties. Generalised information on hammer characteristics is oftenused in the analysis. Alternatively, site measurements may be made to determine .theboundary conditions and calibrate the results for the hammer system and particular geologicalconditions. Essentially, this approach allows an estimate of pile bearing capacity and anevaluation of pile driveability, and assists in hammer selection. Typical solutions from awave equation analysis are in the form of capacity versus penetration resistance curves, andpredicted maximum stresses in the pile.

It should be noted that the rated hammer energy can differ substantially from actualperformance, and some soil parameters pertaining to wave equation analysis are 'model-dependent' empirical values and may not be measured directly. The analysis should only beused to provide general guidance on likely pile behaviour.

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5.4.4 Use of Soil Mechanics Principles

(1) General. The ultimate bearing capacity of a pile may be assessed using soilmechanics principles. The capacity may be assumed to be the sum of skin friction and baseresistance.

(2) Critical depth concept. The shaft and end-bearing resistance in a uniform soilmay generally be expected to be directly proportional to vertical effective stress. Based onmodel tests on piles in granular materials, Vesic (1977) suggested that beyond a critical depththere will be little increase in both skin friction and base resistance. This is supported byBolton (1986) who predicted that the base capacity would tend towards a limiting value ifvariations in the angle of shearing resistance (</>') with confining pressure are allowed for.

However, Kulhawy (1984) concluded from theoretical considerations that the skinfriction and base resistance do not reach a limit at the so-called critical depth. In examiningthe available test results, Kraft (1991) considered that there are no data from full-scale fieldtests that provide conclusive evidence of limiting values for unit skin friction and baseresistance. However, he found that the rate of increase in resistance, especially basecapacity, appears to decrease with increasing depth in a homogeneous sand. Similarly, Altaeeet al (1992a & b) and Fellenius & Altaee (1995) concluded from analysis of instrumentedpiles that the critical depth concept is not valid when corrections are made for residualstresses in the piles. On the other hand, Kraft (1990) suggested that the critical depth conceptmay be applicable for calcareous sands which are prone to crushing due to pile driving.

In view of the conflicting views in the literature on limiting shaft and base resistance,further research on this aspect is required before a conclusion can be drawn. Limitedavailable data from instrumented piles in saprolites in Hong Kong indicate, a complex andsomewhat erratic distribution of 'local' skin friction. For practical purposes, the methodsdescribed below may be adopted without the need to make specific allowance for criticaldepth effects for skin friction.

(3) Bored piles in granular soils. Based on plasticity theories, the ultimate endbearing stress, qb for piles in granular soils may be expressed in terms of vertical effectivestress, av', and the bearing capacity factor, Nq as

qb = N qa vf ...... ............. (5.2)

Nq is generally related to the angle of shearing resistance, <£'. Values of Nq factorquoted in the literature vary considerably. Davies & Chan (1981) suggested the valuespresented by Brinch Hansen (1970), while both Poulos & Davis (1980) and Fleming et al(1992) recommended the use of factors derived by Berezantzev et al (1961) which is alsosupported by Vesic (1967). Poulos and Davis (1980) further suggested that for thedetermination of Nq, the value of <£' should be reduced by 3° to allow for possible looseningeffect of installation. For general design purposes, it is suggested that the Nq values proposedby Berezantzev et al (1961) as presented in Figure 5 may be used.

The calculated ultimate base stress should conservatively be limited to 10 MPa, unlesshigher values have been justified by load tests. It is prudent to apply an upper limit on theqb value because the angle of friction and hence the base resistance may be reduced due to

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suppressed dilation and possible crushing of soil grains at high pressure.

The ultimate shaft friction (rs) for piles in granular soils may be expressed in termsof effective stresses as follows :

rs = c' + Ks avf tan <5S ^^

rs = ]8 avf (when c' is assumed to be zero) (5.4)

where Ks = coefficient of horizontal pressure which depends on the relative density andstate of the soil, method of pile installation, and material, length and shape ofthe pile

av' = mean vertical effective stress<5S = angle of friction along pile/soil interface|S = shaft friction coefficient

The angle of interface friction is primarily a function of the nature of pile material andthe state of the ground, and it can be reasonably determined in a shear box test (Lehane,1992). Typical ranges of <5S values are shown in Table 6. Ks may be related to thecoefficient of earth pressure at rest, K0 (see Table 7). However, the determination of K0 isnotoriously difficult as it is a function of stress history and not a fundamental soil property.In the case of saprolites, the KQ value may be lower than that given by the conventional

formula K0 = 1 - sin <£' due to possible effects of bonding (Vaughan & Kwan, 1984). Thisis supported by deduction from field measurements in Hong Kong as reported by Endicott(1982) and Howat (1985).

It should be noted that the Ks value is a function of the method of pile construction.In view of the uncertainties associated with assessing K0 and the effects of constructionmethod, it may be more reasonable to consider the combined effect as reflected by the (3values deduced from load tests on piles in saprolite. It must be noted that in relating rs toav' with the use of the (3 factor, it is assumed that there is no cohesion component (c'),Although there may be some cohesion for undisturbed saprolite, the effect of construction oncf of the soil at the interface with the pile is difficult to evaluate and may be variable. The(3 values back-analyzed from pile load tests would have included any contribution from cf inthe measured rs.

So (1991) postulated that the skin friction of a pile in a bonded soil such as densesaprolite may be dominated by the increase in horizontal stresses due to its tendency to dilateduring shearing. This may explain isolated load test results (e.g. Holt et al, 1982; Sweeney& Ho, 1982) which indicated a continual increase in skin friction at large relativedisplacement of up to about 4% of pile diameter (viz. 39 mm). The importance of thecontribution of skin friction capacity by the increase in horizontal stresses due to dilation isalso highlighted by Nicola & Randolph (1993) and Lehane & Jardine (1994). Based on cavityexpansion theory, So (1991) suggested that the dilation and hence the skin friction in asmaller diameter pile will be greater than that in a large diameter pile. At present, thisremains a conceptual model and has not been sufficiently validated by load test results.However, it is possible that this dilation effect compensates the small insitu stresses in thesaprolite such that pile capacity is broadly similar to that in a sedimentary granular deposit.

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Davies & Chan (1981) gave typical 0 values for bored piles in sand (Table 8). Thesevalues are comparable to those suggested by Meyerhof (1976) for bored piles in granular soils(Figure 6). These may be used for bored piles in sands of sedimentary nature.

Available instrumented load test data from large-diameter bored piles in saprolite(Appendix A) indicate that substantial skin friction is mobilised at a relative pile-soilmovement of about 1 % pile diameter (about 10 to 15 mm), in many cases. The data furthersuggest that the back-analyzed ./3 values at an average relative settlement of 1% pile diameterare generally near the lower bound values recommended by Davies & Chan (1981). Basedon the available load test results in Hong Kong, it is suggested that the calculated averageultimate unit skin friction should be limited to 140 kPa for granitic saprolite unless a highervalue can be justified by site-specific load tests. For preliminary design of piles in saprolite,recommendations given in Tables 6, 7 & 8 may be used to calculate the skin friction usingthe effective stress method. It should be noted that values of Ks and j3 in Tables 7 and 8respectively, are based on back analysis of field test data. Therefore, the effective stressmethod is essentially a semi-empirical design approach.

It should be cautioned that data also exist in Hong Kong for large-diameter bored pilesshowing very low skin friction in dense to very dense granitic saprolite, although it is possiblethat these were a result of problems associated with pile construction. In view of the possibleadverse effects of construction, the assumptions concerning design parameters, constructionmethod and workmanship should be verified by load testing of instrumented piles whenfriction bored piles are proposed, until sufficient local experience has been built up.

The behaviour of piles in colluvium may be greatly affected by the presence ofboulders (e.g. Chung & Hui, 1990). However, a lower bound estimate may be made basedon the properties of the matrix material and using the effective stress method for design.

(4) Driven piles in granular soils. The concepts presented for the calculation of end-bearing capacity and skin friction for bored piles in granular soils also apply to driven pilesin granular soils. The main difference lies in the choice of design parameters which shouldreflect the pile-soil system involving effects of densification and increase in horizontal stressesin the ground due to pile driving. McNicholl et al (1989b) stated that limited load tests ondriven piles in Hong Kong suggested that the qb values can range from 16 MPa to over21 MPa.

For end-bearing capacity calculation, the Nq values given in Figure 5 can be used.Kishida (1967) suggested that for the determination of Nq, the value of </>' can be taken as theaverage of the </>' value prior to driving and 40°, to allow for the influence on & due to piledriving. The calculated ultimate base stress should be limited to 15 MPa (Tomlinson, 1994).Higher values can be used if supported by pile load tests.

Methods have been put forward by Fleming et al (1992) and Randolph et al (1994) toaccount for the dependence of <£' on stress level in the determination of end bearingresistance. Fleming .etal's method, which involves an iterative procedure, relates <£' to therelative density of soil corresponding to the mean effective stress at failure at pile toe level,and critical state friction angle, c/>cv '. It should be cautioned that this approach involvesgeneralization of the stress dilation behaviour of granular material. Experience of applyingthis approach to pile design in Hong Kong is limited.

53

The results of load tests on driven piles in granular soils are subject to considerablescatter, generally more so than for bored piles (Meyerhof, 1976). There is a range ofproposed design methods relating /3 values to <£' which can give very different results. Fordriven piles in sands of sedimentary nature, the average recommendations of Davies & Chan(1981) may be adopted.

Available load test results in weathered rocks in Hong Kong suggest that the /? valuesare near the lower bound of the range of values suggested by Davies & Chan (1981). Verylimited data are available for driven piles in dense saprolite (Appendix A). For driven pilesin saprolite, the design may be carried out using Tables 6, 7 & 8, having regard to the typeof pile, consistency of material, previous experience, etc.

In the event that the pile founded within a competent stratum but is within ten pilediameters from a weak stratum (either above or below the founding stratum), the calculatedultimate end bearing capacity should be adjusted according to the procedure put forward byMeyerhof (1976; 1986).

(5) Bored piles in clays. The shaft resistance of bored piles in clays develops rapidlywith pile settlement and is generally fully mobilized when the pile settlement is about 0.5percent of the pile diameter. On the contrary, the base resistance is fully mobilized only untilthe pile settlement amounts to 10 to 20 percent of the base diameter (Whitaker and Cooke,1966).

The ultimate end bearing stress for piles in clays is often related to the undrained shearstrength, cu, as follows :

Qb = Nc cu (5.5)

where Nc may generally be taken as 9 when the location of the pile base below the groundsurface exceeds four times the pile diameter. For shallower piles, the Nc factor may bedetermined following Skempton (1951).

The ultimate skin friction (rs) of piles in stiff over-consolidated clays can be estimatedbased on the semi-empirical method as follows :

rs = a cu (5.6)

where a is the adhesion factor. Based on back analyses of load tests on instrumented boredpiles, Whitaker & Cooke (1966) reported that the a value lies in the range of 0.3 to 0.6,while Tomlinson (1994) and Reese & O'Neill (1988) reported a values in the range of 0.4to 0.9. In the above correlations, the cu is generally determined from unconsolidatedundrained triaxial compression tests. The effects of sample size on cu are discussed by Patel(1992).

The above design method suffers from the shortcoming that cu is dependent on the testmethod and size of specimens. Caution should be exercised in extrapolating beyond thebounds of the database.

Burland (1973) suggested that an effective stress analysis is more appropriate for piles

54

in stiff clays as the rate of pore-pressure dissipation is so rapid that for normal rates of loadapplication, drained conditions generally prevail in the soil adjacent to the pile shaft. Burland& Twine (1989) re-examined the results of a large number of tests on bored piles in over-consolidated clays and concluded that the shaft friction in terms of effective stress correspondsto angles of friction which are at or close to the residual angle of shearing resistance (<£/).The value of skin friction for piles in an over-consolidated clay may therefore be estimatedfrom the following expression :

rs = Ks av' tan <£r ' (5.7)

where Ks can be assumed to be K0.

The above is also supported by instrumented load test results reported by O'Riordan(1982).

Both the undrained and effective stress methods can generally be used for the designof piles in clays. The use of the undrained method relies on an adequate local database oftest results. In the case where piles are subject to significant variations in stress levels afterinstallation (e.g. excavation, rise in groundwater table), the use of the effective stress methodis recommended, taking due account of the effects on the Ks values due to the stress changes.

(6) Driven piles in clays. Based on available instrumented pile test results, a designcurve is put forward by Nowacki et al (1992) as shown in Figure 7.

Field studies of instrumented model piles carried out to investigate the fundamentalbehaviour of driven cylindrical steel piles in stiff to very stiff clays (e.g. Coop & Wroth,1989; Lehane, 1992) indicated that a residual shear surface is formed along or near the shaftof a pile during installation. Bond & Jardine (1991) found the shear surfaces to bediscontinuous when the pile is driven or jacked into the ground rapidly but to be continuouswhen the jacking is carried out slowly. The observed instrumented model pile behaviour hasbeen summarised by Nowacki et al (1992). Thus, the design should be checked by usingequation 5.6.

(7) Other factors affecting skin friction. Comparative studies of the ultimate shaftresistance of bored piles installed with or without bentonite slurry in granular and cohesivesoils have been carried out (e.g. Touma & Reese, 1974; Fleming & Sliwinski, 1977; Majanoet al, 1994). These studies showed no significant difference in performance with the twomethods of installation. Tomlinson (1994) suggested that the skin friction on bored pilesconstructed using bentonite slurry be reduced by 10% to 30% for prudence. Experience withlarge diameter bored piles and barrettes in saprolite in Hong Kong indicates that the use ofbentonite slurry may not produce detrimental effects on pile performance, provided itsproperties are strictly controlled. Caution concerning piles involving the use of bentoniteslurry which indicate very low unit skin friction as noted in Section.5.4.4(3) above shouldhowever be noted.

The skin friction may also be affected by the concrete fluidity and pressure (Van Impe,1991). The method and speed of casting, together with the quality of the concrete(water/cement ratio and consistency), may have a profound effect on the horizontal stressesand hence the shaft friction that can be mobilised. Denial and Reese (1983) reported that

55

unless the slump of concrete is at least 175 mm and the rate of placement is at least 12 m perhour and a concrete mix with small-size aggregates is used, the pressures exerted by the fluidconcrete will be less than the hydrostatic pressure, which can result in lower skin frictionparticularly in soils with high K0 values.

(8) Effect of soil plug in open-ended pipe piles. For open-ended steel tubes,consideration will need to be given to assessing whether the pile will act in a plugged modeor unplugged mode.

When subject to working load, an open-ended pile with soil plug does not behave inthe same way as a closed-ended pile driven to the same depth. This is because in the formercase, the soil around and beneath the open end is not displaced and compressed to the sameextent as that beneath a closed-ended pipe. Tonilinson (1994) suggested that for open-endedpipe piles driven in cohesive materials, the ultimate bearing capacity can be taken as the sumof the skin friction along the external perimeter of the shaft and the ultimate base resistance,i.e. ignoring the internal skin friction between soil plug and pile. The skin friction andultimate base resistance can be determined as if the pile were closed-ended, but a reductionfactor of 0.8 and 0.5 respectively should be applied. The base resistance should be calculatedusing the gross cross-sectional area of the pile.

The size of soil plug in a pipe pile driven into granular soil is very limited. Theultimate bearing capacity of the pile can be taken as the sum of the external and internal shaftfriction and the base resistance on the net cross-sectional area of the pile toe; or the end-bearing resistance of the plug, whichever is less (American Petroleum Institute, 1987).Tomlinson (1994), based on field observations, suggested that the base resistance of open-ended pipe piles should be limited to 5 MPa irrespective of the diameter of the pile or of thedensity of the soil into which they are driven. This limiting value should be used inconjunction with a safety factor of 2.5.

5.4.5 Correlation with Standard Penetration Tests

(1) General. Semi-empirical correlations have been developed relating both skinfriction and end-bearing capacity of piles founded in granular soils to N values. Such aprocedure would provide an approximate means of allowing for variability of the strata acrossa site in normalising and extrapolating the results of load tests. In most of the correlationsthat have been established, the N values generally refer to uncorrected values before pileinstallation.

Because of the varying degree of weathering of the parent rock in Hong Kong, thelocal practice is that SPT is often continued to much higher N values than in most othercountries (Brand & Phillipson, 1984). However, the carrying out of SPT to very high valuesmay damage the shoe which can subsequently lead to erroneous results. The guidance givenin Geoguide 2 : Guide to Site Investigation (GCO, 1987) concerning termination of the testin very dense soils should be followed.

A rule-of-thumb method for use in the design of caissons and bored piles has been inuse in Hong Kong for some years (Chan, 1981). This method is based on the criterion thatan allowable base bearing pressure of 1.2 MPa is associated with an SPT N value of 180 for

56

'dry' condition and an N value of 240 'below the water table'. The allowable skin frictionis either ignored or limited to 10 kPa without the need to be justified by a load test.However, as discussed by Malone (1987), the rule-of-thumb generally results in unrealisticdistribution of mobilised resistance and gross over-design of large-diameter bored pilesfounded in saprolite. Similarly, Lumb (1983) showed, on the basis of his interpretation ofpile tests in Hong Kong that significant skin friction can be developed in granitic saprolite.

(2) End bearing. Meyerhof (1986) proposed that the ultimate end bearing capacityof piles in granular soils may be estimated from average SPT N values near pile pointcorrected for an effective overburden pressure of 100 kPa using the approximate relationshipshown in Figure 8 for both driven and bored piles in granular soils. For piles exceeding0.5 m base diameter, an empirical reduction factor, rb, should be used (Figure 8).

Malone et al (1992) analyzed the results of instrumented load tests carried out onlarge- diameter bored piles and barrettes embedded in saprolites in Hong Kong. They foundthat the resistance (in kPa) mobilised at the base of the pile at a settlement corresponding to1 % pile diameter is in the range of 6 to 13 times the uncorrected average N values at the baseof the pile.

(3) Skin friction. For piles driven into granular soils, results presented by Meyerhof(1976 & 1986) are scattered and indicate that the average skin friction generally ranges from1.5-N! to 7 N! (kPa), for SPT Nl values up to about 30, where Nl is the mean SPT N valuecorrected for an effective overburden pressure of 100 kPa. He suggested that the averageskin friction may be taken as 2 Nj (kPa) for driven piles in sand. Based on very limited datain Hong Kong, the skin friction for small-displacement piles such as steel H-piles can betaken as 1.5 N to 2 N (kPa) for design, for an N value up to about 50. N is the uncorrectedmean SPT value.

Based on observations of load tests on precast prestressed concrete piles in HongKong, Ng (1989) proposed that rs in the range of 4 N to 7 N (kPa) may be taken for designin saprolite with a limiting average skin friction of 250 kPa. This is generally consistent withthe 'rule-of-thumbf adopted in Hong Kong that rs = 4.8 N (Siu & Kwan, 1982) for N valuesup to about 60 for driven piles. It is recommended that the relationship of rs = 4.5 N (kPa)may be used for design of large-displacement driven piles in saprolite.

For bored piles in sand, load test results summarised by Meyejhof (1976) indicate thatthe average ultimate skin friction generally ranges from 1 N to 2.5 N (kPa), for N values upto 60. For design, Meyerhof (1986) suggested that rs should be taken as 1 Nj (kPa), Similarrecommendations were put forward by Shioi & Fukui (1982) for bored piles in sand in Japan.Wright & Reese (1979) found that the ratio of local skin friction to local N values generallyranges from 1.5 to 5 (kPa) for bored piles in sand. Overall, it can be seen that therelationship of rs = 1 N to 1.5 N (kPa) may be used for bored piles in granular soils.

For saprolite in Hong Kong, instrumented load tests on large diameter bored piles andbarrettes (Appendix A) suggest that the ratio of the average mobilised skin friction to N valuegenerally ranges between 0.8 and 1.4 (kPa) at an average relative pile/soil settlement of about1% pile diameter. It is found that the skin friction is, in some cases, practically fullymobilised at about this relative movement. The mobilised skin friction was found to dependlargely on the construction method and workmanship, as well as the geology and undisturbed

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ground conditions. Compared to bored piles in other tropically weathered soils, it appearsthat the above observed ratio of rs/N is low. For instance, Chang & Broms (1991) reporteda ratio of rs/N ranging from about 0.7 to 4 (kPa) for bored piles in residual soils andweathered rocks in Singapore for N values up to 60, and suggested the relationship of rs/Nof 2 (kPa) for design purposes. This is also supported by Ho (1993) for piles in weatheredgranite in Singapore for N values up to 75. The discrepancy may be due to differences ingeology, and methods of interpretation.

For preliminary design of large-diameter bored piles, barrettes and hand-dug caissonsin sandy granitic saprolite below sea level in Hong Kong, the relationship of rs/N of 0.8 to1.4_(kPa) may be used, with N value limited to 200. Very limited data suggest the ratio ofrs/N may be lower in volcanic saprolite (Appendix A).

The design method involving correlations with SPT results is empirical in nature, andthe level of confidence is not high particularly where the scatter in SPT N values is large.If load tests on preliminary piles are not carried out, this design approach should be checkedusing the effective stress method based on soil mechanics principles (Section 5.4.4(3)), andthe smaller calculated capacity adopted for design.

In traditional design of small-diameter bored piles involving pressure grouting orpressurising the concrete in Hong Kong, the empirical relationship of rs = 4.8 N to 5 N(kPa), ignoring the contribution from the base, is generally used for N values up to about 40,usually with a factor of safety of 3 (Chan, 1981). Lui_et al (1993) reported a design of postgrouted mini-piles based on the relationship of rs = 5 N (kPa), where N is limited to 100 andthe factor of safety is taken to be 3, which has been satisfactorily verified by instrumentedpile load tests.

5.4.6 Correlation with Other Insitu Tests

Piles may be designed based on correlations with other types of insitu tests such ascone penetration tests, pressuremeter tests and dilatometer tests.

Cone penetration tests are best suited for silts and sands that are loose to mediumdense (such as hydraulically-placed fill and alluvial sands) but may meet premature refusalin dense sands and gravels. The test is generally unsuitable in weathered rocks.

Semi-empirical methods have been developed relating results of Static ConePenetration Tests (i.e. Dutch Cone or piezocones) to the bearing capacity of piles (e.g.Bustamante&Gianeselli, 1982; Meyerhof, 1986; Tomlinson, 1994; Lehane & Jardine, 1994).

In Hong Kong, pressuremeter (e.g. Menard Pressuremeter) has occasionally been usedto measure the deformation characteristics and limit pressure values of granitic saprolite forthe design of foundations (Chiang & Ho, 1980). Baguelin et al (1978) presented curvesrelating ultimate skin friction and base capacity to the pressuremeter limit pressure, for bothdriven and cast-insitu piles. These may be used for a rough preliminary assessment but, dueto lack of a reliable local database, they should be confirmed by load tests.

Dilatometers may be used to provide an index for a number of properties including the

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insitu horizontal stress. These indices may, in principle, be used to correlate with pilecapacity.

The use of correlations developed overseas based on insitu tests for Hong Kongconditions should be done with caution as a number of other factors may also influence thepile capacity, e.g. different geological formations (Tomlinson, 1994).

5.5 AXIALLY LOADED PILES IN ROCK

5.5.1 General

For the purpose of pile design in Hong King, rock is generally taken to be fresh tomoderately decomposed rock or partially weathered rock having a rock content greaterthan 50%. For a short rigid pile founded on top of rock surface, it is acceptable to neglectthe insignificant adhesion along its sides in the soil layers and assume that the applied loadis transferred to the base. For piles socketed in rock, the shaft resistance of the rock socketcould be significant and should be taken into account in the design (Section 5.5.4). Wherethe rock surface is sloping, the lowest point intersected by the pile should be conservativelytaken as the start of the rock socket.

For the ideal model of a long pile constructed through soil and founded on rock, thedegree of load transfer in the portion of the pile shaft embedded in soil will depend on theamount of relative movement arising from base deflection and elastic compression of theshaft, i.e. it will be a function of the relative shaft and base stiffness. In a corestone-bearingweathering profile, the distribution of load in the pile is likely to be complex and may behighly variable.

The settlement of piles founded on rock which have been designed on the basis ofbearing capacity theories should always be checked as this is generally the governing factorin, for example, weak rocks, closely-fractured rocks and moderately to highly decomposedrocks.

In the past the capacity of concrete piles on rock was generally limited by the strengthof the concrete. With the use of high strength concrete, the capacity of piles on rock maynow be controlled by the strength as well as the compressibility of the rock mass which needsto be assessed more accurately.

5.5.2 Driven Piles on Rock

Where the joints are widely-spaced and closed, very high loads can be sustained bythe rock mass and the design is unlikely to be governed by bearing capacity of the ground.In such ground conditions, piles driven to refusal can be designed based on permissiblestructural stresses of the pile section. Where the joints are open or. clay-filled, the rock massbelow the pile tip may yield under load/The assessment of the load deformation propertiesof such rock mass can be made using the empirical relationships developed by Barton (1986)(see Section 5.13.2(5)).

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5.5.3 Bored Piles on Rock

(1) General The methods of designing bored piles founded on rock may be broadlyclassified under four categories :

(a) semi-empirical methods,

(b) bearing capacity theories,

(c) the use of presumptive bearing values from codes, and

(d) insitu tests.

(2) Semi-empirical methods. Peck et al (1974) suggested a semi-empirical correlationbetween allowable bearing pressure and Rock Quality Designation (RQD) as shown inFigure 9. The correlation is intended for a rock mass with discontinuities "that are tight orare not open wider than a fraction of an inch; settlement of the foundation should not exceedhalf an inch". The use of such correlation should only be regarded as a crude first step inrock foundation design (Peck, 1976). It should be noted that RQD may be biased dependingon the orientation of the boreholes in relation to the dominant discontinuities.

An alternative semi-empirical method of assessing the allowable bearing pressure ofpiles founded in a rock mass has been proposed in the Canadian Foundation EngineeringManual (Canadian Geotechnical Society, 1992). This method, described in Figure 10,assumes that the allowable bearing pressure is equal to the product of the average unconfinedcompressive strength and modification factors which account for spacing and aperture ofdiscontinuities in the rock mass, width of the foundation and effect of socket depth (Ladanyi& Roy, 1971).

In practice, it is suggested that both semi-empirical methods should be used todetermine the allowable bearing pressure, and the lower bound value adopted for design.

(3) Bearing capacity theories. Sowers (1979) proposed that the failure modes shownin Figure 11 should be considered in design. For a thick rigid layer overlying a weaker one,failure can be by flexure, with the flexural strength being approximately twice the tensilestrength of the rock. For a thin rigid layer overlying a weak one, failure can be by punching,i.e. tensile failure of the rock mass. For both cases, bearing failure of the underlying weaklayer should be checked. Failure in a rock mass with open joints is likely to occur by uniaxialcompression of the rock columns. For rock mass with closed joints, a general wedge shearzone will develop. Where the rock mass is widely-jointed, failure occurs by splitting of therock beneath the foundation which eventually leads to a general shear failure. Reference maybe made to Figure 11 for foundation design using bearing capacity theories.

The relevant strength parameters (c1 and 0 f ) may be estimated on the basis of a semi-empirical failure criterion such as the modified Hoek & Brown criterion (Hoek et al, 1992).

Kulhawy & Carter (1992a) developed a lower bound bearing capacity solution forfoundations on rock in terms of the Hoek & Brown's (1980) strength criterion for jointedrock mass. The strength parameters for the rock mass can be empirically determined from

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CSIR rock mass rating RMR (Bieniawski, 1974) or NGFs rock mass quality index Q (Bartonet al, 1974). Bieniawski & Orr (1976) proposed that the RMR values can be adjusted toaccount for the effect of joint orientation on the load capacity and settlement of thefoundation.

In a widely-jointed strong rock, the allowable load on the pile will be governed by thepermissible structural stresses of the pile section. In principle, the use of very high strengthconcrete ranging from, say, 60 to 75 MPa (Kwan, 1993) will increase the allowable pilecapacity. However, there may be practical problems associated with achieving such highconcrete strength given the requirements for high workability for self compaction of pilingconcrete, and possible concrete placement by means of tremie under a stabilising fluid. Otherpotential problems, such as thermal effects and creep, will also need to be considered.Sufficient field trials, including testing of cores of the pile, will be required to prove thefeasibility of very high strength concrete for piling.

(4) Presumptive bearing values. As an alternative to using rational design methods,foundations for structures which are not unduly sensitive to settlement may be designed usingpresumptive bearing values given in a local code. Irfan & Powell (1991) summarized thepresumptive allowable bearing stress for rock as specified by various Building Codes. InHong Kong, presumptive bearing values for granitic and volcanic rocks ranging from 3 MPato 7.5 MPa have customarily been used (Table 9). These presumptive bearing values reflectlocal experience and may be used in routine work or for preliminary designs. Account shouldbe taken of nearby excavation and/or orientation of discontinuities, together with theinteraction effects of adjacent piles at different elevations in the case of rock with a slopingsurface.

The design of piles founded on decomposed granitic rock in Hong Kong based onpresumptive bearing values has been verified by instrumented load tests (Koutsoftas & Reese,1989). A load-tested pile on granitic rock, described as having been weathered excessively,settled considerably when loaded to about 6 700 kPa. However, the settlement in subsequentloading cycle was essentially elastic suggesting that the prestressing due to load cycling duringthe load test had increased the base stiffness.

A design based on presumptive bearing pressures, which are generally on the safe side,may not be the most cost effective. The use of presumptive values should not be a substitutefor consideration of settlement, particularly if the structure is susceptible to foundationmovements.

The use of the percentage total core recovery as the sole means of determiningfounding level in rock could be misleading because the value can be affected by theeffectiveness of the drilling technique used in retrieving the core. Similarly, the use of RQDas the sole means of determining founding level can also lead to erroneous results becauseit does not take into account the discontinuity, aperture, continuity or condition, or thepresence and character of any infilling material. Irfan & Powell (1985) concluded that theuse of a rock mass weathering classification system, in conjunction with simple index tests,will be superior to the use of RQD or total core recovery alone, and can enable limitedengineering data to be applied successfully over a large site area.

In the case of a belled pile founded on rock, consideration will need to be given to

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assessing whether there will be sufficient shear capacity in the concrete to transmit the forcescorresponding to mobilisation of the allowable pressure over the whole area of the bell.

(5) Insitu tests. The load-deformation characteristics of the base of a rock foundationmay be evaluated by insitu tests such as plate loading tests, Goodman Jack, pressuremeter orfull-scale load tests.

5.5.4 Rock Sockets

The side resistance of a rock socket is significantly affected by the roughness of theinterface (Seidel & Haberfield, 1994). There is a wealth of local load test results on rockanchors which justify the conventional assumption in Hong Kong of an allowable shaftfriction of 0.5 to 1 MPa. The lower end of the range of skin friction applies to Grade IIIrock while the upper end applies to Grade II or better rock. However, there is a lack ofinformation on large-diameter bored piles acting as rock sockets, and no systematic study hasbeen carried out to investigate the effect of differences in shaft roughness for rock socketsconstructed by machine drilling and hand-dug caisson techniques.

Assuming full contact between the pile and the rock, the load distribution in a rocksocket is primarily a function of its geometry, and the relative stiffness of concrete and therock mass. As a first approximation, the load on the pile may be apportioned between endbearing and shear due to bond in accordance with Figure 12 following Pells & Turner (1979).Kulhawy & Goodman (1980) put forward a design approach which allows for bond raptureand considers post rupture behaviour. This latter approach is generally only relevant wheresignificant pile movement can be tolerated.

Using data from a large number of load tests on rock-socketed piles, Horvath et al(1983) established an empirical relationship between unit bond strength and compressivestrength of the weaker materials, i.e. the weaker of concrete and rock (Figure 13).

A range of methods has been proposed in the literature for designing rock sockets(Irfan & Powell, 1991). Where a particular design method predicts a much higher calculatedcapacity than that given by Horvath et al (1983), the design value should be justified by asufficient number of load tests.

Full-scale instrumented load tests on rock sockets in Hong Kong are scarce. Lam etal (1991) reported that the average skin friction mobilised in a 1 m diameter hand-dug caissonsocketed in moderately to slightly decomposed granite was 670 kPa. For this pile, thecorresponding unconfined compressive strength of the concrete cores was only about 7 MPa.Shiu & Chung (1994) reported the results of load tests on two 219 mm diameter mini-pilessocketed in moderately decomposed granite. The average compressive strength of the groutwas 45 MPa and a maximum local skin friction of 1.75 MPa was mobilised. Both of theabove test results compare favourably with the upper-bound design line given in Figure 13.

For preliminary design of rock sockets in a widely-jointed rock, it is suggested thatthe average relationship given in Figure 13 can be used. However, as there is a generalpaucity of data to validate the design approach, and given the potential influence ofconstruction method and workmanship on skin friction, it is advisable to carry out a load test

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to verify the design parameters.

For piles socketed into rock, the safety margin against ultimate bearing failure of theground is likely to be large, and should not control design. However, it is difficult to cleanthe base of a rock socket, and debris may be left behind. In view of this, it is prudent toestimate the allowable working load using a minimum mobilisation factor of 1.5 on theultimate shaft resistance calculated from Figure 13, ignoring the contribution from the base,so as to limit the settlement at working load. Alternatively, for short large-diameter boredpiles, the base of the rock socket can be preloaded to allow bearing pressure to be mobilizedat small settlements (Horvath et al, 1983).

5.6 UPLIFT CAPACITY OF PILES

5.6,1 Piles in Soil

Some published test results (e.g. Radhakrishnan & Adams, 1973; Broms & Silberman,1964) indicate that the uplift resistance in skin friction is less than the corresponding skinfriction in compression, possibly by up to 50% less in a granular soil. Fellenius (1989)suggested that this may be due to the influence of the reduction in vertical effective stress inthe ground and Poisson's ratio effect under tension loading.

Fleming et al (1992) considered that the interpretation of many pile load tests tookinsufficient account of the residual stresses which existed after pile installation, andconsequently the end bearing capacity of the pile was under-estimated and the skin frictionover-estimated. They suggested that there is no systematic difference in the skin frictionwhich may be mobilised by an unstressed pile loaded either in tension or compression.

Premchitt et al (1988) observed that the pattern of residual stresses developed after piledriving was complex and erratic. Therefore, it is difficult to generalise for design purposes.It was noted by Premchitt et al that the residual skin friction and base resistance locked inafter pile driving were not associated with well-defined displacements or an applied loading.Furthermore, the consideration of the skin friction associated with the applied loading in aload test (i.e. zeroing the instrumentation immediately prior to a load test) represents thecondition of actual working piles supporting superstructure loads. With driven piles, anumber of researchers have also emphasized the importance of the dependence of radialhorizontal stresses and skin friction on the relative position of the pile tip as the pile isadvanced, based on observations made in instrumented piles (e.g. Lehane, 1992;Lehane et al, 1993).

For design purposes, it is • recommended that the skin friction of bored piles undertension may be calculated in the same way as for skin friction for compression piles(Sections 5.4.4(3) & 5.4.4(5)). For driven piles, in view of the uncertainties associated withthe distribution of residual stresses after driving and the available capacity having alreadybeen partially mobilised, it is recommended that the skin friction under tension be takenconservatively as 75% of that under compression (Section 5.4.4(4) & Section 5.4.4(6)),unless higher values can be justified by a sufficient number of load tests.

For relatively slender piles, such as mini-piles, the tendency for the pile to contract

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radially under tension may lead to lower radial stresses and hence reduced skin friction.Fleming et al (1992) estimated that this reduction may amount to 10% to 20%.

Any possible suction effects that may develop at the base of a pile should bedisregarded for prudence as this may not be reliable.

The working load under tension loading, Qwt is given by the following :

Qwt = (Qs/Fs) + Wp' (5.8)

where Qs = ultimate skin friction capacity under tensionFs = factor of safety

Wpf = effective self weight of the pile

It is recommended that a minimum factor of safety of 2.5 to 3.5 (Table 4) should beprovided on the ultimate skin friction in tension.

For piles with an enlarged base, Dickin & Leung (1990) reviewed existing designmethods and investigated the uplift behaviour of such piles embedded in sand using acentrifuge (Figure 14). For dense sand, they found sensible agreement with earlier researchon anchor plates and published field data. It was concluded that the best prediction for pilecapacity in dense sand when compared with the centrifuge test results is that given byVermeer & Sutjiadi (1985). For loose sand, the existing methods appear to over-predict theultimate resistance to uplift with the exception of the simple vertical slip surface modelproposed by Majer (1955). In the absence of relevant field data from instrumented piles, itis suggested that the above recommendations may be adopted for preliminary design.However, given that the design methods are based on model test results, they should beconfirmed by a pull out test.

The potential problems associated with the construction of bell-out are discussed inSection 7.3.4(12).

5.6.2 Rock Sockets

The uplift behaviour of large-diameter rock sockets has not been investigated in detailand a pragmatic approach may be adopted for routine design (Kulhawy & Goodman, 1987).The cone failure mode of rock mass is normally the governing criterion under pull out. Itis suggested that a reduction of 30% of the socket bond capacity in compression may beassumed in view of the lack of a proper design basis and possible negative Poisson's ratioeffect, unless a higher value can be justified by load tests.

Bonding at the base of the socket will be governed by the tensile strength of theweaker of the rock or concrete. However, given the potential construction problems due todifficulties in achieving proper base cleanliness, possible intermixing of tremie concrete andwater and bentonite, etc, it is suggested that this should be conservatively ignored in design.

Rock anchors are sometimes provided for tension piles to increase their uplift capacity.The uplift resistance of the rock anchors depends on the permissible stress in the anchor,

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bond strength between the anchor, the grout, and the rock, and the weight of rock mass andoverlying soil lifted by the anchor or a group of anchors (Tomlinson, 1994).

The actual shape of the mass of rock lifted depends on the degree of jointing, fissuringand the inclination of the bedding planes of the rock. For a heavily jointed or shattered rock,a cone with a half angle of 30° will give a conservative estimate for the pull-out resistance(Tomlinson, 1994). Shear at the interface between the cone surface and the surrounding rockshould be neglected.

For rock mass with steeply inclined joint sets, the weight of the rock cone should beconservatively assessed.

5.6.3 Cyclic Loading

Cyclic loading leads to at least three aspects of soil response that are not encounteredunder static loading conditions (Poulos, 1989b), namely :

(a) degradation of pile-soil resistance,

(b) loading rate effects, and

(c) accumulation of permanent displacements.

Detailed studies using full-scale instrumented piles (e.g. Ove Arup & Partners, 1986;Karlsrud & Nadim, 1992) suggest that the reduction in the static capacity is much greater intwo-way type cyclic loading (i.e. load reversed between tension and compression) comparedto one-way cyclic loading (i.e. both maximum and minimum loads applied in the same senseor direction). A useful review of piles in granular soils subjected to cyclic loading is givenby Poulos (1989b) and Turner & Kulhawy (1990). Jardine (1992) summarised the state-of-the-art on pile behaviour in clays under cyclic loading.

5.7 LATERAL LOAD CAPACITY OF PILES

5.7.1 Vertical Piles in Soil

The lateral load capacity of a pile may be limited in three ways :

(a) shear capacity of the soil,

(b) structural (i.e. bending moment and shear) capacity of thepile section, and

(c) excessive deformation of the pile.

For piles subject to lateral loading, the failure mechanisms of short piles under lateralloads as compared to those of long piles differ, and different design methods are appropriate.The stiffness factors as defined in Figure 15 will determine whether a pile behaves as a rigid

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unit (i.e. short pile) or as a flexible member (i.e. long pile).

As the surface soil layer can be subject to disturbance, suitable allowance should bemade in the design, e.g. the resistance of the upper part of the soil may be ignored asappropriate.

Brinch Hansen (1961) proposed a method of calculating the ultimate lateral resistanceof a c'-<£! material, which can be used for short rigid piles (Figure 16).

Methods of calculating the ultimate lateral soil resistance for fixed-headed and free-headed piles in granular soils and clays are put forward by Broms (1964a & b). The theoryis similar to that of Brinch Hansen except that some simplifications are made in respect ofthe distribution of ultimate soil resistance with depth. The design methods for short and longpiles in granular soils are summarised in Figures 17 and 18 respectively.

Broms1 methods have been extended by Poulos (1985) to consider the lateral loadcapacity of a pile in a layered soil.

The design approaches presented above are simplified representations of the pilebehaviour. Nevertheless, they form a useful framework for obtaining a rough estimate of thelikely capacity, and experience suggests that they are generally adequate for routine design.Where the design is likely to be governed by lateral load behaviour, load tests should becarried out to justify the design approach and verify the design parameters.

The bending moment and shearing force in a pile subject to lateral loading may beassessed using the method by Matlock & Reese (1960) as given in Figures 19 and 20. Thetabulated values of Matlock & Reese have been summarised by Elson (1984) for easyreference. This method models the pile as an elastic beam embedded in a homogeneous, ornon-homogeneous soil. The structural capacity of a long flexible pile is likely to govern theultimate capacity of a laterally-loaded pile.

For relatively short (less than critical length given in Section 5.13,3(3)) end-bearingpiles, e.g. piles founded on rock, with toe being effectively fixed against both translation androtation, they can be modelled as a cantilever encastred at the bottom and either fixed or freeat the top depending on restraints on pile head. The lateral stiffness of the overburden canbe represented by springs with appropriate stiffness.

The minimum factors of safety recommended for design are summarised in Table 4.The design of a vertical pile to resist lateral load is usually governed by limiting lateraldeflection requirements.

For piles in sloping ground, the ultimate lateral resistance can be affected significantlyif the piles are positioned within a distance of about five to seven pile diameters from theslope crest. Based on full-scale test results, Bhushan et.al (1979) proposed that the lateralresistance for level ground be factored by (1/(1 + tan 0S)), where 0S is the slope angle.Alternatively, Siu (1992) proposed a simplifying method for determining the lateral resistanceof a pile in sloping ground taking into account three-dimensional effects.

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5.7.2 Inclined Loads

If a vertical pile is subjected to an inclined and eccentric load, the ultimate bearingcapacity in the direction of the applied load is intermediate between that of a lateral load anda vertical load because the passive earth pressure is increased and the vertical bearingcapacity is decreased by the inclination and eccentricity of the load. Based on model tests,Meyerhof (1986) suggested that the vertical component Qv of the ultimate eccentric andinclined load can be expressed in terms of a reduction factor rf on the ultimate concentricvertical load Q0, as given in Figure 21.

The lateral load capacity can be estimated following the methods given inSection 5.7.1. Piles, subjected to inclined loads, should be checked against possible buckling(Section 5.12.4), pile head deflection (Section 5.13.3) and induced bending moments.

5.7.3 Raking Piles in Soil

A common method of resisting lateral loads is to use raking piles. For the normalrange of inclination of raking piles used in practice, the raking pile may be considered as anequivalent vertical pile subjected to inclined loading.

Comments on the method of determining the applied load on raking piles are given inSection 6.5.3.

5.7.4 Rock Sockets

Based on elastic analyses, Poulos (1972) has shown that a rock socket constructedthrough soil has little influence on the lateral behaviour under working loading unless the pileis relatively stiff (i.e. with a stiffness factor, Kr, of greater than 0.01, see Section 5.13.3).For such stiff piles, e.g. large-diameter bored piles, the contribution of the socket to thelateral load capacity may be accounted for using the principles presented by Poulos & Davis(1980) assuming a distribution of ultimate lateral resistance mobilised in the rock. Where therock level dips steeply, consideration should be given to assuming different ultimate resistancein front of and behind the pile.

In a heavily-jointed rock mass with no dominant adversely-orientated joints, a wedgetype analysis may be carried out using c', & values determined based on the modified Hoek& Brown criterion (Hoek et al, 1992). Alternatively, Carter & Kulhawy (1992) presenteda theoretical method for determining the lateral load capacity of a pile socketed in a rockmass, based on considering a long cylindrical cavity in an elasto-plastic, cohesive-frictional,dilatant material. In assessing the ultimate lateral resistance, due consideration must be givento the rock mass properties including the nature, orientation, spacing, roughness, aperturesize, infilling and groundwater conditions of discontinuities.

The possibility of a joint-controlled failure mechanism should be checked (GEO,1993). Joint strength parameters reported in Hong Kong have been summarised by Brand etal (1983). Alternatively, the rock joint model presented by Barton et al (1985) may be used.

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5.7.5 Cyclic Loading

Cyclic or repeated loading may lead to problems of degradation of soil resistance andstiffness, or 'post-holing' where gaps may form near the ground surface. Long et al (1992)reviewed the methods of analysing cyclic loading on piles in clays. Reference may be madeto Poulos (1988a) for the design of piles in granular soils subjected to cyclic loading.

5.8 NEGATIVE SKIN FRICTION

5.8.1 General

Piles installed through compressible materials (e.g. fill or marine clay) can experiencenegative skin friction. This occurs on the part of the shaft along which the downwardmovement of the surrounding soil exceeds the settlement of the pile. Negative skin frictioncould result from consolidation of a soft deposit caused by dewatering or the placement offill. The dissipation of excess pore water pressure arising from pile driving in soft clay canalso result in consolidation of the clay.

The magnitude of negative skin friction that can be transferred to a pile depends on(Bjerrum, 1973) :

(a) pile material,

(b) method of pile construction,

(c) nature of soil, and

(d) amount and rate of relative movement between the soil andthe pile.

In determining the amount of negative skin friction, it would be necessary to estimatethe position of the neutral point, i.e. the level where the settlement of the pile equals thesettlement of the surrounding ground. For end-bearing piles, the neutral point will be locatedclose to the base of the compressible stratum.

5.8.2 Calculation of Negative Skin Friction

For friction piles, various methods of estimating the position of the neutral point, bydetermining the point of intersection of pile axial displacement and the settlement profile ofthe surrounding soil, have been suggested by a number of authors (e.g. Fellenius, 1984).However, the axial displacement at the pile base is generally difficult to predict without pileload tests in which the base and shaft responses have been measured separately. As a roughguide, NAVFAC (1982) recommended that the ratio of the depth of the neutral point to thelength of the pile within a uniform settling stratum may be taken as 0.75, which is reasonableif the ground settlements are not substantial. Where the induced ground settlements are large,the neutral point may be near the base of the compressible layer, particularly if the pile is notsubject to structural loading (e.g. Lee & Lumb, 1982).

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The mobilised negative skin friction, being dependent on the horizontal stresses in theground, will be affected by the type of pile. For steel H-piles, it is important to check thepotential negative skin friction with respect to both the total surface area and thecircumscribed area relative to the available resistance (Broms, 1979).

The effective stress or (3 method (Section 5.4.4) may be used to estimate the magnitudeof negative skin friction on single piles (Bjerrum et al, 1969; Burland & Starke, 1994). Fordesign purposes, the range of (3 values given in Figure 22 may be used for assessing thenegative skin friction.

Various methods of calculating the negative skin friction based on elastic continuumassumptions have also been proposed. By assuming that the surrounding soil behaves as anelastic homogeneous material with a constant modulus of elasticity which is relevant for clays,Davis & Poulos (1969) used the equations of Mindlin (1936) to calculate the stressdistribution along a pile. The negative skin friction is determined using an influence factorIn and an approximate assessment of the effect of delayed installation of piles relative tocommencement of ground settlement can be made.

In general, it is only necessary to take into account negative skin friction incombination with dead loads and sustained live load, without consideration of transient liveload. Transient live loads will usually be carried by positive skin friction, since a very smalldisplacement is enough to change the direction of the skin friction from negative to positive,and the elastic compression of the piles alone is normally sufficient. In the event where thetransient live loads are larger than twice the negative skin friction, the critical load conditionwill be given by (dead load + sustained live load + transient live load). The aboverecommendations are based on consideration of the mechanics of load transfer down a pile(Broms, 1979) and the research findings (Bjerrum et al, 1969; Fellenius, 1972) that verysmall relative movement will be required to build up and relieve negative skin friction, andelastic compression of piles associated with the transient live load will usually be sufficientto relieve the negative skin friction. Caution needs to be exercised however in the case ofshort stubby piles founded on rock where the elastic compression may be insufficient to fullyrelieve the negative skin friction. In general, the customary local assumption of designingfor the load combination of (dead load + full live load + negative skin friction) is on theconservative side.

If the ground settlement profile is known with reasonable certainty, due allowance maybe made for the portion of the pile shaft over which the relative movement is insufficient tofully mobilise the negative skin friction (i.e. movement less than 0.5% to 1% of pilediameter). However, the reduction in negative skin friction is normally not very significantand may be conservatively ignored.

The way safety factors are incorporated in the design of piles subjected to negativeskin friction requires careful consideration. In the case of a pile embedded in a compressiblematerial, the problem of negative skin friction may be regarded as a settlement problem andthe overall factor of safety against ultimate failure remains unchanged (Fellenius, 1989).However, in the case of a pile founded on rock, it becomes a bearing or structural capacityproblem (Canadian Geotechnical Society, 1992).

If the pile settlement under working load is small, e.g. short large-diameter piles end-

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bearing on rock, there may not be significant relief of negative skin friction due to pilesettlement and the following method suggested by NAVFAC (1982) should be used :

Qall = (Qult/F s)-Pn - ' ' • (5.9)

where Qau = the allowable pile loadQuit = the ultimate resistance below the neutral point

Fs = global factor of safetyPn = ultimate negative skin friction

It should be noted that negative skin friction can have the dual effect of inducingdowndrag and reducing the available capacity as a result of reduction in overburden pressure.

Tomlinson (1994) considered it over-conservative to subtract the full negative skinfriction from the working load on the pile in order to arrive at the allowable load. Hesuggested that the required total ultimate pile load can be obtained by multiplying the workingload only by the normal safety factor, and then checking that the safety factor given by theultimate load divided by the sum of working load and negative skin friction is still a'reasonable' value. The approach suggested by Tomlinson (1994) is not considered rationalparticularly in the case where Pn is large and is close to, or exceeds, Qan.

Where the pile settlement is sufficient to relieve a proportion of the negative skinfriction, Broms (1981) proposed the following approach :

Qall = (Quit - Pn)/Fs (5-1Q)

In this case, the problem becomes one of assessing whether the working loadsettlement will be acceptable and the factor of safety should be assessed as appropriate tosatisfy serviceability criteria. This approach does not ensure that the pile settlement willstabilize if the surrounding soil continues to settle.

Poulos (1990b) demonstrated how pile settlement can be determined using elastictheory with due allowance for yielding condition at the pile/soil interface.

5.8.3 Field Observations in Hong Kong

Lee & Lumb (1982) reported the results of an instrumented closed-ended tubular pileloaded by a 2 m high embankment for about a year. The back-analyzed /3 values fordowrndrag in the fill/marine sand and in the marine clay were about 0.61 and 0.21,respectively, which are broadly consistent with the recommended values given in Figure 22.

Available long-term monitoring data on piles driven into saprolite (i.e. friction piles)through an old reclamation (i.e. fill placed more than 20 years ago) indicates that nosignificant negative skin friction builds up in the long term after building occupation (Ho &Mak, 1994). This is consistent with the fact that primary consolidation under the reclamationfill is complete, and that no significant settlement and negative skin friction will result unlesslarge reductions in the water level are imposed (Lumb, 1962), or soft clays with a potentialfor developing large secondary consolidation settlement are present.

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5.8.4 Means of Reducing Negative Skin Friction

Possible measures that can be adopted to reduce negative skin friction include coatingwith bitumen or asphalt, together with use of an enlarged point or collar at the bottom of thepile, and use of sacrificial protection piles around the structure, and various groundimprovement techniques such as electro-osmosis (Broms, 1979).

Field tests carried out by Lee & Lumb (1982) for a site in Tuen Mun indicate thatcoating of steel tubular piles can be effective in reducing negative skin friction, In this case,load tests demonstrated that downdrag with coating was only 14% of that with no coating.

Steel tubular piles which are protected with an inner coating of 2 mm thick bitumen,and an outer protective coating of polyethylene plastic of minimum thickness 3.5 mm werealso reported to have been effective in reducing negative skin friction when driven throughreclaimed land in Japan (Fukuya et al, 1982).

In Norwegian practice, a minimum bitumen coating of 1 mm is used for steel piles and2 mm for concrete piles (Simons & Menzies, 1977).

The effectiveness of any slip coating will depend on the extent of damage sustainedduring pile handling and driving and should be confirmed by site trials. The durability of thecoating must also be considered as bitumen has been observed to be attacked bybacteriological action in marine clays (Simons & Menzies, 1977).

5.9 TORSION

It is rarely necessary to design piles for torsion loading. Reference may be made toRandolph (1981b) for piles subject to torsion.

5.10 PRELIMINARY PILES FOR DESIGN EVALUATION

The best way to determine pile behaviour is to carry out full-scale load tests onrepresentative preliminary piles to obtain suitable parameters to verify the design assumptions.It would be necessary to characterize the ground conditions so as to permit generalisation andextrapolation of the test results to other areas of the site. The need for preliminary pilesshould be carefully assessed by the designer, having regard to familiarisation with the groundconditions, the type of pile, previous experience and the scale of the project.

The preliminary piles should preferably be load tested to the ultimate state or at leastto sufficient movements beyond those at working conditions. The use of internalinstrumentation will provide valuable information on the load transfer mechanism and willfacilitate back analysis. Instrumented piles should be considered particularly in unfamiliaror difficult ground conditions and when novel pile types are being proposed. Load testingof preliminary piles can enhance the reliability of the design and can, in some cases, lead toconsiderable savings.

Where possible, the preliminary piles should be located in the area with the most

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adverse ground conditions. They should be constructed in the same manner using the sameplant and equipment as for working piles so as to evaluate the adequacy of workmanship andthe method of construction. It is recommended that an exploratory borehole be sunk at orin the vicinity of the preliminary pile position for retrieving undisturbed samples andappropriate insitu tests in order to characterize the ground conditions and facilitate back-analysis of test results.

The number of preliminary piles should be selected on the basis of a range ofconsiderations including :

(a) ground conditions and their variability across the site,

(b) type of pile and method of construction,

(c) previous documented evidence of the performance of thesame type of pile in similar ground conditions,

(d) total number of piles in the project, and

(e) contractor's experience.

As a rough guide, it is recommended that at least two preliminary piles for the first100 piles (with a minimum of one preliminary pile for smaller contracts) should be loadtested when there is a lack of relevant experience (e.g. in unfamiliar ground conditions or useof novel pile types). Where the pile performance is particularly prone to the adequacy ofquality control and method of construction (e.g. large-diameter bored piles in saprolite), atleast one preliminary pile should be load tested for the first 100 piles. In both instances,where a contract involves a large number of piles when the total number of piles exceeds 200,the number of additional preliminary piles may be based on the frequency of one per every200 piles after the first 100 piles.

If any of the preliminary piles fail the load test marginally, the pile capacity shouldbe downgraded as appropriate. However, if the piles fail the test badly and the failure isunlikely to be due to over-optimistic design assumptions, the reasons for the failure shouldbe investigated in detail. The number of piles to be further tested should be carefullyconsidered.

For large-diameter bored piles or barrettes, it may be impractical to carry out a loadtest on a full size preliminary pile. Load tests on a smaller diameter preliminary pile maybe considered, provided that :

(a) it is constructed in exactly the same way as piles to be usedfor the foundation, and

(b) it is instrumented to determine the shaft and base resistanceseparately.

Details of pile instrumentation and interpretation of load tests are covered inChapter 8.

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5.11 PILE DESIGN IN KARST MARBLE

The design of piles founded in karst marble requires consideration of the karstmorphology, loading intensity and layout of load bearing elements. The main problemaffecting the design is the presence of overhangs and cavities which may or may not beinfilled. The stability of the piled foundation will depend on the particular geometry of suchkarst features, and the rock mass properties, particularly of the discontinuities.

For large-diameter bored piles with concentrated loads, the 'angle of stress dispersion'concept is sometimes used. This concept requires that there should be no major cavitieswithin a zone below the pile base as defined by a cone of a given angle to the vertical, withinwhich sensible increase in vertical stress would be confined. This approach is acceptable asan aid to judgement in pile design. Careful consideration should be given to the nature andextent of the adverse karst features and of their positions, in plan and elevation, in relationto nearby piles and to the foundation as a whole, together with the quality of the interveningrock.

It must be stressed that no simple design rules exist which could overcome all thepotential problems associated with karst formation. To allow for uncertainties arising fromthe karst morphology, suitable redundancy should be included in the pile design, whicheverpiling system is used, to allow for the effects of dissolution features below pile toe level andfor possible pile damage sustained during driving. No definite guidelines can be given forthe percentage of redundancy as this depends on the extent, nature and geological backgroundof the karst features and the type of pile. Each site must be considered on its own merits.Some discussion on the consideration of redundancy factors (i.e. the factor by which the pilecapacity is downrated) is given by Chan (1994b). Where redundant piles are provided forpossible load redistribution, the effect of this possible re-distribution should be considered inthe design of the pile cap. Where the foundation consists of a number of pile caps rather thanthe usual single raft, it may be necessary to increase the redundancy, and to ensure adequateload transfer capacity between the units by means of inter-connecting ground beams.

For local areas with adverse karst features, it may be feasible to 'bridge over' theproblem area provided the outline of the area is well defined by the ground investigation.

Foundations on karst marble in Yuen Long and Ma On Shan areas have successfullybeen constructed using bored piles founded beneath cavities, and steel H-piles withredundancy factors applied and hard driven to refusal into the surface karst of marbleformation. Small-diameter cast-in-place piles 'floating1 in the soil strata well above the topof marble surface have also been used.

Chan et al (1994) proposed a system for classifying the marble rock mass in HongKong. A new index termed Marble Quality Designation (MQD) is put forward. This indexis a combined measure of the degree of dissolution voids, and the physical and mechanicalimplications of fractures or a cavity-affected rock mass (Figure 23). The marble rock massis classified in terms of MQD values. This marble rock mass classification system is usedin the interpretation of the karst morphology, and offers a useful means for site zoning interms of the degree of difficulties involved in the design and construction of foundations. Asummary of the proposed classification system, together with comments on its engineeringsignificance, is given in Table 10. An approach to the design of piles on karst marble in

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Hong Kong, which makes use of the new classification system, is described by Ho et al(1994).

McNicholl et al (1989b) reported the presence of a weak, structureless soil layer abovethe marble rock surface in the Tin Shui Wai area and suggested that this might have beenaffected by slumping and movement of fines into the underlying cavities. Mitchell (1985)reported similar findings in Malaysia. The significance of this weaker material on the piledesign should be carefully considered.

5.12 STRUCTURAL DESIGN OF PILES

5.12.1 General

Structural design of piles should be carried out in accordance with the requirementsin local structural codes and regulations. The piles should be capable of withstanding both thestresses induced during handling and installation as well as during their service life.

The structural forces that will arise in service will be very much a function of therelative stiffnesses of the three components, viz. the structure, piles and soil.

It is suggested that the analysis of structural forces in a soil-structure interactionproblem at serviceability conditions may be based on characteristic loads, i.e. without beingfactored, together with best estimates of soil properties (Section 2.9). The calculatedstructural forces should be regarded as working load effects and should be compared with theallowable working stress of the material or factored up in accordance with appropriate limitstate structural codes to obtain ultimate-load effects for structural design as appropriate.

5.12.2 Lifting Stresses

The adequacy of reinforcement in precast reinforced (including prestressed) concretepiles to resist bending should be checked for the case of bending stresses induced by lifting.The possibility of the pile being rotated about its longitudinal axis during lifting should alsobe taken into account.

5.12.3 Driving and Working Stresses

The stresses induced in a pile during driving may be calculated using a wave equationanalysis (Section 5.4.3). The maximum driving stresses must not exceed the acceptablelimiting stresses on the pile material.

An alternative and simplified approach, which is commonly adopted, is to limit theworking stresses under static loading such that hard driving is not required to achieve thepenetration resistance necessary for the calculated ultimate bearing capacity. Many codeslimit the working structural stresses which can be carried by a pile. In Hong Kong, thelimiting average compressive stresses on the nominal cross-sectional area at working loada r e : ' , " • • . : " • • . ' ; ' . : . • • • ' . ' • • • . ' ' " . ' • . . ' ' ' • • • ' • ' : ' ^ , \'

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(a) precast reinforced concrete piles : 0.4 fc/Fd,

(b) steel piles : 0.6 fy/Fd

(c) cast-in-place concrete piles :

(i) The appropriate limitations of design stresses of theconcrete in the case of concreting in dry conditions

(ii) 80% of the appropriate limitations of design stressesof the concrete, in the case where groundwater islikely to be encountered during concreting.

where fc is the specified grade strength of concrete, fy is yield stress of the steel and Fd isthe design safety factor on driving, with a minimum value of 2 for dead and imposed loads,and 1.6 for combined dead, imposed and wind loads (Hong Kong Government, 1993).

More guidance on precautions to be taken during construction is given in Section7.2.5(2).

5.12.4 Bending and Buckling of Piles

H-piles and steel tubular piles are flexible and may deflect appreciably from theintended alignment during driving. Specifications normally allow tolerances in alignment andplan position at cut-off level, e.g. 1 in 75 deviation from vertical and 75 mm deviation inplan for vertical piles. A method of calculating the bending stresses caused by eccentricloading is explained in Figure 24. In general, pile buckling should be checked assuming thepile is at maximum allowable tolerance in alignment and plan. In situations where there aresignificant horizontal loads (and/or moments) applied at pile head, the combined effectsshould be considered in pile design.

Piles rarely buckle except for long slender piles (e.g. mini-piles) in very soft groundor where piles have been installed through significant cavities in karstic marble. Studies onthis problem have been carried out by a number of researchers (e.g. Davisson &Robinson, 1965; Reddy & Valsangkar, 1970). Analyses indicate that buckling will beconfined to the critical length of the pile under lateral loading (Figure 25).

5.12.5 ^Mini-piles

In Hong Kong, the allowable structural capacity of a mini-pile has generally beenassessed conservatively by ignoring the contribution of the grout even under compression.The allowable stress of the steel will be that given by local structural codes or buildingregulations. It would be more rational, and in line with overseas practice, to make a suitablycautious allowance for the contribution by the grout. Available instrumented pile tests (Luiet'al, 1993) indicated that the grout did contribute to the load-carrying capacity.

Provided that strict site control and testing of the grouting operation (Section 7.3.5(3))are implemented, the design strength of the grout may be taken notionally as 75% of the

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measured characteristic cube strength, with an allowable stress of 25% of the design strengthcontributing to the allowable structural capacity of the pile.

Where very high strength steel bars (e.g. Dywidag bars) are used, care should be takento consider the effect of strain compatibility between the steel and the grout, as the availablestrength of the steel may not be mobilised due to failure of the grout.

5.13 DEFORMATION OF SINGLE PILES

5.13.1 General

Various analytical techniques have been developed to predict pile deflections. Thesetechniques provide a convenient framework for deriving semi-empirical correlations betweenequivalent stiffness parameters back-analyzed from load tests and index properties of theground. Some of the analytical methods can also be extended to evaluate pile interactioneffects in an approximate manner, thus enabling an assessment of pile group behaviour to bemade within the same framework.

5.13.2 Axial Loading

(1) General. The various approaches that have been proposed for predicting pilesettlement can be broadly classified into three categories :

(a) load transfer methods,

(b) elastic continuum methods, and

(c) numerical methods.

In calculating movements, the stiffness of the founding materials at the appropriatestress level needs to be determined. For normal pile working loads (of the order of 40% to50% of ultimate capacity), Poulos (1989a) has shown that the non-linear nature of soilbehaviour generally does not have a significant effect on the load-settlement relationship forsingle piles.

(2) Load transfer method. In the load transfer method proposed by Coyle & Reese(1966) for piles in soil, the pile is idealised as a series of elastic discrete elements and the soilis modelled by elasto-plastic springs. The load-displacement relationship at the pile head,together with the distribution of load and displacement down the pile, can be calculated usinga stage-by-stage approach as summarised in Figure 26.

The axial load transfer curves, sometimes referred to as 't-z' curves, for the springsmay be developed from theoretical considerations. In practice, however, the best approachto derive the load transfer curves is by back analysis of an instrumented pile test because thistakes into account effects of pile construction .

The load transfer method provides a consistent framework for considering the loadtransfer mechanism and the load-deformation characteristics of a single pile.

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(3) Elastic continuum methods. The elastic continuum method, sometimes referredto as the integral equation method, is based on the solutions of Mindlin (1936) for a pointload acting in an elastic half-space. Different formulations based on varying assumptions ofshear stress distribution along the shaft may be used to derive elastic solutions for piles.Solutions using a simplified boundary element method formulation are summarised by Poulos& Davis (1980) in design chart format.

In the method by Poulos & Davis (1980), the pile head settlement (6t) of anincompressible pile embedded in a homogeneous, linear elastic, semi-infinite soil mass isexpressed as follows :

6t = P I S * /E S D ---- ............. (5.11)

where P — applied vertical loadIs* = influence factor for pile settlementEg = Young's modulus of founding materialD = pile diameter

The pile settlement is a function of the slenderness ratio (i.e. pile length/diameter,L/D), and the pile stiffness factor, K, which is defined as follows :

K = EpRA/Es .................. (5.12)

where Ep = Young's modulus of the pileRA = ratio of pile area Ap to area bounded by outer circumference of pile

sk

Influence factor, Ip , can be applied to allow for the mode of load transfer (i.e. frictionor end-bearing piles), effects of non-homogeneity, Poisson's ratio, pile compressibility, pile-soil slip, pile base enlargement, nature of pile cap, etc. Reference should be made to Poulos& Davis (1980) for the appropriate values.

The ratio of short term (immediate) settlement to long term (total) settlement can bededuced from elastic continuum solutions. For a single pile, this ratio is typically about 0.85to 0.9 (Poulos & Davis, 1980).

In a layered soil where the modulus variation between successive layers is not large,the modulus may be taken as the weighted mean value (Eav) along the length of the pile (L)as follows :

E = 1/L E C E d ...... .......... <5"13)av

where Ej = modulus of layer idj = thickness of layer in = number of different soil layers along the pile length

An alternative formulation also based on the assumption of an elastic continuum wasput forward by Randolph & Wroth (1978). This approach uses simplifying assumptions onthe mode of load transfer and stress distribution to derive an approximate closed-formsolution for the settlement of a compressible pile (Figure 27). A method of dealing with a

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layered soil profile based on this approach is given by Fleming et al (1992).

It should be noted that the above elasticity solutions are derived assuming the soil isinitially unstressed. Thus, pile installation effects are not considered explicitly except in thejudicious choice of the Young's modulus. Alternative simplified elastic methods have beenproposed by Vesic (1977) and Poulos (1989a) including empirical coefficients for driven andbored piles respectively in a range of soils. Similar approximate methods may be used fora preliminary assessment of single pile settlement provided a sufficient local database of pileperformance is available.

For piles founded on rock, the settlement at the surface of the rock mass can becalculated by the following formula assuming a homogeneous elastic half space below the piletip :

where d^ = settlement at the surface of the rock massq = bearing pressure on the rock mass

Cd = depth correction factorCs = shape and rigidity correction factorj>r = Poisson's ratio of the rock mass

Db = pile base diameterEr = Young's modulus of rock mass

The depth correction factor may be obtained from Figure 28, which has beenreproduced from Burland & Lord (1970). The shape and rigidity factor is shown in Table 11(Perloff, 1975).

For piles founded in a jointed rock, Kulhawy & Carter (1992a & b) have also putforward a simplified method for calculating the settlements.

(4) Numerical methods. Fleming (1992) developed a method to analyse and predictload-deformation behaviour of a single pile using two hyperbolic functions to describe theshaft and base performance individually under maintained loading. These hyperbolicfunctions are combined with the elastic shortening of the pile. By a method of simplelinkage, based on the fact that the hyperbolic functions require only definition of their origin,their asymptote and either their initial slope or a single point on the function, elastic soilproperties and ultimate loads may be used to describe the load-deformation behaviour of thepile.

The load-deformation behaviour of a pile can also be examined using numericalmethods including rigorous boundary element analyses (e.g. Butterfield & Bannerjee, 1971a& b) or finite element analyses (e.g. Randolph, 1980; Jardine et al, 1986). Distinct elementmethods (e.g. Cundall, 1980) may be appropriate for piles in a jointed rock mass.

These numerical tools are generally complicated and time consuming, and are rarelyjustified for routine design purposes, particularly for single piles. The most useful applicationof numerical methods is for parametric studies and the checking of approximate elasticity

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solutions.

An application of the finite element method is reported by Pells & Turner (1979) forthe solution derivation and design chart compilation for the settlement of rock-socketed pilesbased on linear elastic assumptions. This work has been extended by Rowe & Armitage(1987a & b) to consider effects of pile-soil slip on the settlement. More recent work has beenreported by Kulhawy & Carter (1992a & b). Gross approximations would have beennecessary if this boundary value problem were to be solved by the integral equation method.

It is considered the above simplified design charts may reasonably be used for detaileddesign purposes.

(5) Determination of deformation parameters. A useful review of the assessment ofsoil stiffness is given by Wroth et al (1979). In principle, the stiffness can be determinedusing a range of methods including directly from insitu tests (e.g. plate tests, pressuremeters,etc) or indirectly from insitu tests based on empirical correlations (e.g. SPT, CPT, etc),surface geophysical methods using Rayleigh waves (Clayton et al, 1993), back analysis ofinstrumented prototype structures, etc.

The general practice in Hong Kong has been to obtain stiffness parameters forsaprolites using correlations with SPT N values. Table 12 summarises the correlationsreported in the literature for weathered granite in Hong Kong.

The stiffness of the soil under the action of a pile will be dependent on the pileinstallation method and workmanship, and stress level. For preliminary design of bored pilesfounded in saprolites, the following correlation may be used in the absence of any site-specific data :

EV' = 0.8 N to 1.2 N (MPa) (5.15)

where Ey' is'the drained vertical Young's modulus of the soil, and N is the uncorrected SPTvalue.

Vesic (1969) suggested that the stiffness for a driven pile system in sands may be takento be approximately four times that for a corresponding bored pile system.

Based on available load test results in Hong Kong, the following correlation may beused for preliminary analysis of driven piles in granitic saprolite :

Ey' =.3.5 N to 5.5 N (MPa) (5.16)

Densification during pile driving will lead to an increase in soil stiffness but the effectmay be variable and site dependent. Limited data in Hong Kong have shown that the iy/Nfratio may be in the order of about 2.5 to 3 where Nf is the SPT blowcount after pile driving.

In determining the relevant rock mass deformation parameters, consideration shouldbe given to influence of non-homogeneity, anisotropy and scale effects. Deformation of arock mass is often governed by the characteristics of discontinuities. There are a number ofmethods that can be used to assess the deformation properties including :

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(a) relating the deformation modulus of the rock mass to thedeformation modulus of the intact rock (the latter can becorrelated to the uniaxial compressive strength, ac) bymeans of a mass factor denoted as 'j! factor (BSI, 1986),

(b) empirical correlations with the CSIR Rock Mass Rating,RMR (Bieniawski, 1978; Serafim & Pereira, 1983), and

(c) semi-empirical relationships with properties of the rockjoints (Barton, 1986), which can be used in complexcomputer codes based on distinct element models of therock mass (Cundall, 1980).

In Barton's model, the surface roughness, shear and dilation behaviour of a rock jointis represented by semi-empirical relationships, which are characterized by the properties ofthe joint and are also functions of the normal stress and displacement at the joint. Theparameters required by the model can be determined in the laboratory using tilt tests, Schmidthammer tests and simple rock joint profiling techniques.

For practical design, an estimate of the order of magnitude of rock mass deformationis adequate as a sensitivity check. The elastic continuum method is widely used and isgenerally adequate for routine design problems in assessing the pile head settlement atworking conditions. The appropriate deformation parameters should be derived using morethan one assessment method or be obtained directly from load tests.

5.13.3 Lateral Loading

(1) General. The response of piles to lateral loading is sensitive to soil propertiesnear the ground surface. As the surface layers may be subject to disturbance, reasonablyconservative soil parameters should be adopted in the prediction of pile deflection. Anapproximate assessment of the effects of soil layering can be made by reference to the workby Davisson & Gill (1963) or Pise (1982).

Poulos (1972) studied the behaviour of a laterally-loaded pile socketed in rock. Heconcluded that -socketing of a pile has little influence on the horizontal deflection at workingload unless the pile is sufficiently rigid, with a stiffness factor under lateral loading, Kr,greater than 0.01, where Kr = Ep Ip/(Es L4), and Ip and L are the second moment of areaand length of the pile respectively.

The effect of sloping ground in front of a laterally-loaded pile was analyzed by Poulos(1976) for clayey soils, and by Nakashiffia et al (1985) for granular soils. It was concludedthat the effect on pile deformation will not be significant if the pile is beyond a distance ofabout five to seven pile diameters from the slope crest.

The load-deflection and load-rotation relationships for a laterally-loaded pile aregenerally highly non-linear. Three approaches have been proposed for predicting thebehaviour of a single pile :

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(a) equivalent cantilever method,

(b) subgrade reaction method, and

(c) elastic continuum method.

Alternative methods include numerical methods such as the finite element andboundary element methods as discussed in Section 5.13.2(4). However, these are seldomjustified for routine design problems.

A useful summary of the methods of determining the horizontal soil stiffness is givenby Jamiolkowski & Garassino (1977).

It should be noted that the currently available analytical methods for assessingdeformation of laterally-loaded piles do not consider the contribution of the side shearstiffness. Some allowance may be made for barrettes loaded in the direction of the long sideof the section with the use of additional springs to model the shear stiffness and capacity inthe subgrade reaction approach.

Where the allowable deformation is relatively large, the effects of non-linear bendingbehaviour of the pile section due to progressive yielding and cracking together with its effecton the deflection and bending moment profile should be considered (Kramer & Heavey,1988). The possible non-linear structural behaviour of the section can be determined bymeasuring the response of an upstand above the ground surface in a lateral load test.

(2) Equivalent cantilever method. The equivalent cantilever method is a grosssimplification of the problem and should only be used as an approximate check on the othermore rigorous methods unless the pile is subject to nominal lateral load. In this method, thepile is represented by an equivalent cantilever and the deflection is computed for either free-head or fixed-head conditions. Empirical expressions for the depths to the point of virtualfixity in different ground conditions are summarised by Tomlinson (1994).

The principal shortcoming of this approach is that the relative pile-soil stiffness is notconsidered in a rational framework in determining the point of fixity. Also, the method isnot suited for evaluating profiles of bending moments, etc.

(3) Subgrade reaction method. In the subgrade reaction method, the soil is idealisedas a series of discrete springs down the pile shaft. The continuum nature of the soil is nottaken into account in this formulation.

The characteristic of the soil spring is expressed as follows :

P = kh6h ..Y... . . ; . ; . . . . . . . . . . . (5.17)

Ph = Kh5h . . (5.18)= kh D 6h (for constant Kh)=. % z Sh (for the case of Kh varying linearly with depth)

where p = soil pressure

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kh = coefficient of horizontal subgrade reaction (in units of kN/m3)5h = lateral deflectionPh = soil reaction (in units of kN) per unit length of pileKh = modulus of horizontal subgrade reaction (in units of kPa)D = width or diameter of pilenh = constant of horizontal subgrade reaction (in units of kN/m3), sometimes

referred to as the constant of modulus variation in the literaturez = depth below ground surface

It should be noted that kh is not a fundamental soil parameter as it is influenced by thepile dimensions. In contrast, Kh is more of a fundamental property and is related to theYoung's modulus of the soil, and it is not a function of pile dimensions.

Traditionally, a over-consolidated clay is assumed to have a constant Kh with depthwhereas a normally consolidated clay and a granular soil is assumed to have a Kh increasinglinearly with depth, starting from zero at ground surface.

For a uniform pile with a given bending stiffness (Eplp), there is a critical length (Lc)beyond which the pile behaves under lateral load as if it were infinitely long and can betermed a 'flexible' pile.

The expressions for the critical lengths are given in the following :

Lc = 4(Ep Ip/Kh)1/4 (5.19)

= 4 R for soils with a constant Kh

Lc = 4(Ep Ip/nh)1/5 (5.20)

= 4 T for soils with a Kh increasing linearly with depth

The terms 'R' and T' are referred to as the characteristic lengths by Matlock & Reese(1960) for homogeneous soils and non-homogeneous soils, respectively. They derivedgeneralised solutions for piles in granular soils and clayey soils. The solutions for granularsoils as summarized in Figures 19 and 20 have been widely used in Hong Kong.

A slightly different approach has been proposed by Broms (1964a & b) in which thepile response is related to the parameter /TL, where 18* = (Kh/4EpIp)

1/4, for clays, and to theparameter ??*L, where 77* = (nh/EpIp)

1/5, for granular soils. The solutions for granular soilsare summarised in Figure 29.

In general, the subgrade reaction method can give satisfactory predictions of thedeflection of a single pile provided the subgrade reaction parameters are derived fromestablished correlations or calibrated against similar case histories or load test results.

Typical ranges of values of %, together with recommendations for design approach,are given in Table 13,

The parameter kh can be related to results of pressuremeter tests (CanadianGeotechnical Society, 1992).

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The effects of pile width and shape on the deformation parameters are discussed bySiu (1992).

The solutions by Matlock & Reese (1960) apply for idealised, single layer soil. Thesubgrade reaction method can be extended to include non-linear effects by defining thecomplete load transfer curves. This formulation is more complex and a non-linear analysisgenerally requires the use of computer models similar to those described by Bowles (1988),which can be used to take into account variation of deformation characteristics with depth.In this approach, the pile is represented by a number of segments each supported by a spring,and the spring stiffness can be related to the deformation parameters by empirical correlations(e.g. SPT N values). Due allowance should be made for the strength of the upper, and oftenweaker, soils whose strength may be fully mobilised even at working load condition.

(4) Elastic continuum methods. Solutions for deflection and rotation based on elasticcontinuum assumptions are summarised by Poulos & Davis (1980). Design charts are givenfor different slenderness ratios (L/D) and the dimensionless pile stiffness factors under lateralloading (Kr) for both friction and end-bearing piles. The concept of critical length is howevernot considered in this formulation as pointed out by Elson (1984).

A comparison of these simplified elastic continuum solutions with those of the rigorousboundary element analyses has been carried out by Elson (1984). The comparison suggeststhat the solutions by Poulos & Davis (1980) generally give higher deflections and rotationsat ground surface, particularly for piles in a soil with increasing stiffness with depth.

The elastic analysis has been extended by Poulos & Davis (1980) to account for plasticyielding of soil near ground surface. In this approximate method, the limiting ultimate stresscriteria as proposed by Broms (1965) have been adopted to determine factors for correctionof the basic solution.

An alternative approach is proposed by Randolph (198 la) who fitted empiricalalgebraic expressions to the results of finite element analyses for homogeneous and non-homogeneous linear elastic soils. In this formulation, the critical pile length, Lc (beyondwhich the pile plays no part in the behaviour of the upper part) is defined as follows :

Lc = 2 r0

where r0 = radius of an equivalent circular pileEpe = equivalent Young's modulus of the pile = (EA)p/7r r()

2

Gc = mean value of G* over the critical length, Lc, in a flexible pile(EA)p = cross-sectional rigidity of actual pile

G* = G(l+3?s/4)G = shear modulus of soilys = Poisson's ratio of soil

For a given problem, iterations will be necessary to evaluate the values of Lc and Gc.

Expressions for deflection and rotation at. ground level given by Randolph's elasticcontinuum formulation are summarised in Figure 30.

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Results of horizontal plate tests carried out from within a hand-dug caisson incompletely weathered granite (Whiteside, 1986) indicate the following range of correlation :

Ehf = 0.6 N to 1.9 N (MPa) (5.22)

where Eh' is the drained horizontal Young's modulus of the soil.

The modulus may be nearer the lower bound if disturbance due to pile excavation andstress relief is excessive. The reloading modulus was however found to be two to three timesthe above values.

Data on the appropriate rock mass modulus for piles socketed into rock are limited.Lam et al (1991) reported results of horizontal Goodman Jack tests carried out from withina caisson in moderately to slightly (Grade III/II) weathered granite. The interpreted rock massmodulus was in the range of 3.1 to 8.2 GPa.

In the absence of site-specific field data, the above range of values may be used inpreliminary design of piles subject to lateral loads.

5.14 CORROSION OF PILES

The maximum rate of corrosion of steel piles embedded in undisturbed ground andloaded in compression can be taken to be 0.02 to 0.03 mm/year based on results of researchreported by Romanoff (1962, 1969) and Kinson et al (1981). Moderate to severe corrosionwith a corrosion rate of up to about 0.08 mm/year may occur where piles are driven intodisturbed soils such as fill and reclamation, particularly within the zone of fluctuatinggroundwater level. It should be noted that Romanoff's data suggest that special attentionneeds to be exercised in areas where the pH is below about 4.

Ohsaki (1982) reported the long-term study of over 120 steel piles driven into a varietyof soil conditions and found that the above recommended corrosion rates are generallyconservative.

For maritime conditions, the results of research overseas should be viewed withcaution as the waters in Hong Kong are relatively warm and may contain various pollutantsor anaerobic sulphate-reducing bacteria which greatly increases the risk of pitting corrosion.Faber & Milner (1971) reported fairly extensive underwater corrosion of the foundations toa 40-year old wharf in Hong Kong, involving pitting corrosion of the 3.2 mm thick steelcasing and cavities on the surface of the hearting concrete which required extensiveunderwater repair works.

It is recommended that steel piles above seabed, whether fully immersed, within thetidal or splash zone, or generally above the splash zone, should be fully protected againstcorrosion for the design life (Hong Kong Government, 1992b). This precaution should alsoextend to precast piles where the sections are welded together with the use of steel end plates.

Possible corrosion protection measures that may be adopted include use of copper-bearing or high-yield steel, sacrificial steel thickness, protective paints or coatings (made of

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polyethylene, epoxy or asphalt), together with cathodic protection consisting of sacrificialgalvanic anodes or impressed currents. In a marine environment, steel tubular piles may beinfilled with concrete from pile head level to at least below seabed level and the steel casingabove seabed be regarded as sacrificial. For onshore situations, steel piles may be protectedwith coating or concrete surround within the zone of ground water fluctuation or fill material.The most appropriate measures need to be assessed on a site-by-site basis.

In the case of concrete piles, the best defence against the various possible forms ofattack as summarised by Somerville (1986) is dense, low permeability concrete with sufficientcover to all steel reinforcement. Bartholomew (1980) classified the aggressiveness of the soilconditions and provided guidance on possible protective measures for concrete piles . Somedesigners have followed the recommendations given in BS 8110 (BSI, 1985) in using highstrength concrete of up to about 40 MPa for land piling in order to improve corrosionresistance. However, high strength concrete may not necessarily be dense and homogeneous.Specifying high strength concrete is no guarantee for durability

For concrete piles in maritime conditions, the recommended limits on the propertiesof concrete are as follows (Hong Kong Government, 1992b) :

(a) minimum characteristic strength = 45 MPa,

(b) maximum free water/cement ratio = 0.42, and

(c) concrete cover = 60 mm (for piles fully immersed, andin tidal and splash zones)

= 50 mm (for areas above splash zones)

Criteria (a) and (b) above should apply irrespective of whether the concrete is fullyimmersed, within the tidal or splash zones or located above the splash zone. For concretewithin the tidal and splash zones, crack widths under typical average long term conditionsshould be limited to 0.1 mm. Where protected from direct exposure to the marineatmosphere, reinforced concrete should comply with the recommendations given in BS 8110(BSI, 1985) for 'moderate1 conditions.

With grouted piles such as mini-piles, the minimum cover recommended by theFederation of Piling Specialists (1987) is 30 mm. This may need to be increased incontaminated ground or alternatively a permanent casing may be required.

For piles under permanent tension, the concrete or grout are likely to be cracked underworking conditions and should not be considered as a barrier to corrosion. It is prudent toinclude at least one level of corrosion protection to ensure long-term integrity of the steelelements. The use of sacrificial thickness is permissible, except in aggressive groundconditions.

The presence of leachate and gas in contaminated grounds such as landfills andindustrial areas may pose serious hazards to the construction and functional performance ofpiles (Section 2.6).

In the case where aggregates from potentially reactive rocks or unknown sources are

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used, consideration may need to be given to assessing the susceptibility to alkaline aggregatereaction (AAR), which could lead to severe cracking. Typical approaches include thespecification of the limit on the alkali content (in terms of the total equivalent acid-solublesodium oxide), or the use of 'low alkali' sulphate-resisting cement. The equivalent sodiumoxide content of concrete should not exceed 3 kg per cubic metre to guard against theoccurrence of AAR. Reference should be made to specialist text for guidance on appropriatetests for AAR (e.g. Leung et al, 1995) and how it can best be avoided (see Section 2.6).

ooOs

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6. GROUP EFFECTS

6.1 GENERAL

Piles installed in a group to form a foundation will, when loaded, give rise tointeraction between individual piles as well as between the structure and the piles. The pile-soil-pile interaction arises as a result of overlapping of stress (or strain) fields and could affectboth the capacity and the settlement of the piles. The piled foundation as a whole alsointeracts with the structure by virtue of the difference in stiffness. This foundation-structureinteraction affects the distribution of loads in the piles, together with forces and movementsexperienced by the structure.

The analysis of the behaviour of a pile group is a complex soil-structure interactionproblem. The behaviour of a pile group foundation will be influenced by, inter alia :

(a) method of pile installation, e.g. displacement or replacementpiles,

(b) dominant mode of load transfer, i.e. shaft friction or endbearing,

(c) nature of founding materials,

(d) three-dimensional geometry of the pile group configuration,

(e) presence or otherwise of a ground-bearing cap, and

(f) relative stiffness of the structure, the piles and the ground.

Traditionally, the assessment of group effects is based on some 'rules-of-thumbf orsemi-empirical rules derived from field observations. Recent advances in analytical studieshave enabled more rational design principles to be developed. With improved computingcapabilities, general pile groups with a combination of vertical and raking piles subjected tocomplex loading can be analysed in a fairly rigorous manner and parametric studies can becarried out relatively efficiently and economically.

This Chapter firstly considers the ultimate limit states for a range of design situationsfor pile groups. Methods of assessing the deformation of single piles and pile groups are thenpresented. Finally, some design considerations for soil-structure interaction problems arediscussed.

6.2 MINIMUM SPACING OF PILES

The minimum spacing between piles in a group should be chosen in relation to themethod of pile construction and the mode of load transfer. It is recommended that thefollowing guidelines on minimum pile spacing may be adopted for routine design :

(a) For bored piles which derive their capacities mainly fromshaft friction and for all types of driven piles, minimumcentre-to-centre spacing (s) * the perimeter of the pile(which should be taken as that of the larger pile where pilesof different sizes are used); this spacing should not be lessthan 1 m as stipulated in the Building (Construction)Regulations (Hong Kong Government, 1993).

(b) For bored piles which derive their capacities mainly fromend bearing, minimum clear spacing between the surfaces ofadjacent piles should be based on practical considerations ofpositional and verticality tolerances of piles. It is prudentto provide a nominal minimum clear spacing of about 0.5 mbetween shaft surfaces or edge of bellouts.

The recommended tolerances of installed piles are shown in Table 14 (Hong KongGovernment, 1992a). Closer spacing than that given above may be adopted only when it hasbeen justified by detailed analyses of the effect on the settlement and bearing capacity of thepile group. Particular note should be taken of adjacent piles founded at different levels, inwhich case the effects of the load transfer and soil deformations arising from the piles at ahigher level on those at a lower level need to be examined. The designer should also specifya pile installation sequence within a group that will assure maximum spacing between shaftsbeing installed and those recently concreted.

6.3 ULTIMATE CAPACITY OF PILE GROUPS

6.3.1 General

Traditionally, the ultimate load capacity of a pile group is related to the sum ofultimate capacity of individual piles through a group reduction (or efficiency) factor ??,defined as follows ;

= ultimate load capacity of a pile group /^ ^

sum of ultimate load capacities of individual piles in the group

A number of empirical formulae have been proposed, generally relating the groupreduction factor to the number and spacing of piles. However, most of these formulae giveno more than arbitrary factors in an attempt to limit the potential pile group settlement. Acomparison of a range of formulae made by Chellis (1961) shows a considerable variation inthe values of t\ for a given pile group configuration. There is a lack of sound theoreticalbasis in the rationale and field data in support of the proposed empirical formulae (Fleming& Thorburn, 1983). The use of these formulae to calculate group reduction factors istherefore not recommended.

A more rational approach in assessing pile group capacities is to consider the capacityof both tiie individual piles (with allowance for pile-soil-pile interaction effects) and thecapacity of the group as a block or a row and determine which failure mode is more critical.There must be an adequate factor of safety against the most critical mode of failure.

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The degree of pile-soil-pile interaction which affects pile group capacities is influencedby the method of pile installation, mechanism of load transfer and nature of the foundingmaterials. The group reduction factor may be assessed on the basis of observations made ininstrumented model and field tests as described below. Generally, group interaction does notneed to be considered where the spacing is in excess of about eight pile diameters (CanadianGeotechnical Society, 1992).

6.3.2 Vertical Pile Group in Granular Soils under Compression

(1) Free-standing driven piles. In granular soils, the compacting efforts of piledriving generally result in densification and consequently the group reduction factor may begreater than unity. Lambe & Whitman (1979) warned that for very dense sands, pile drivingcould cause loosening of the soils due to dilatancy and the factor q could be less than unityin this case. This effect is also reflected in the model tests reported by Valsangkar &Meyerhof (1983) for soils with an angle of shearing resistance, <£', greater than 40°.However, this phenomenon is seldom observed in full scale load tests or field monitoring.

Figure 31 shows the findings of model tests on instrumented driven piles reported byVesic (1969). The ultimate shaft capacity of the pile group was observed to have increasedto about three times the capacity of a single pile.

It is generally accepted that, for normal pile spacing, the interaction arising fromoverlapping of stress fields affects only the shaft capacity and is independent of the type ofpile and the nature of the soil. Therefore, it would be more rational to consider groupreduction factors in terms of the shaft friction component only.

The behaviour of a driven pile may be affected by the residual stresses built up duringpile driving. In practice, pile driving in the field could affect the residual stresses of theneighbouring piles to a different extent from that in a model test as a result of scale effectswhich could partially offset the beneficial effects of densification. For design purposes, it isrecommended that a group reduction factor of unity may be taken conservatively fordisplacement piles.

The block or row failure mechanism should also be checked (Figure 32) by consideringthe available shaft friction and end-bearing resistance of the block or row as appropriate.Suitable allowance should be made in assessing the equivalent angle of pile/soil interfacefriction for the portion of failure surface through the Relatively undisturbed ground betweenthe piles.

(2) Free-standing bored piles. Construction of bored piles may cause loosening anddisturbance of granular soils. In practice, the design of single piles generally has madeallowance for the effects of loosening and the problem is therefore to assess the additionaleffect of loosening due to pile group installation. This may be affected to a certain extent bythe initial stresses in the ground but is principally a question of workmanship and constructiontechniques and is therefore difficult to quantify.

Meyerhof (1976) suggested that the group reduction factor could be takenconservatively as 2/3 at customary spacings but no field data were given to substantiate this.

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The results of some load tests on full scale pile groups are summarised by O'Neill (1983),who shows that the lower-bound group reduction factor is 0.7. For design purposes, thegroup reduction factor may be taken as 0.85 for shaft friction and 1.0 for end-bearing,assuming average to good workmanship.

If an individual pile has an adequate margin against failure, there would be no risk ofa block failure of a pile group supported purely by end bearing on a granular soil which isnot underlain by weaker strata. Where the piles are embedded in granular soils (i.e. frictionand end-bearing), both individual pile failure and block failure mechanisms should bechecked.

(3) Pile group with a ground-bearing cap. In the case where there is a ground-bearing cap, the ultimate load capacity of the pile group should be taken as the lesser of thefollowing (Poulos & Davis, 1980) :

(a) Sum of the capacity of the cap (taking the effective area,i.e. areas associated with the piles ignored) and the pilesacting individually. For design purposes, the same groupreduction factors as for piles without a cap may be used.

(b) Sum of the capacity of a block containing the piles and thecapacity of that portion of cap outside the perimeter of theblock.

Care should be exercised in determining the allowable load as the movements requiredto fully mobilise the cap and pile capacities may not be compatible and appropriatemobilisation factors for each component should be used. In addition, the designer shouldcarefully consider the possibility of partial loss of support to the cap as a result of excavationfor utilities, ground settlement, etc.

6.3.3 Vertical Pile Group in Clays under Compression

The extent of installation effects of both driven and bored piles in a clay on pile-soil-pile interaction is generally small compared to that in a granular soil. It should be noted thatthe rate of dissipation of excess pore water pressures set up during driving in clays will beslower in a pile group than around single piles. This may need to be taken into account ifdesign loads are expected to be applied prior to the end of the re-consolidation period.

For a free-standing group of either driven or bored piles, the capacity should be takenas the lesser of the sum of the ultimate capacity of individual piles with allowance for a groupreduction factor and the capacity of the group acting as a block. Reference to the results ofa number of model tests summarised in Figure 33 shows that the group reduction factor forindividual pile failure is generally less than unity and is dependent on the spacing, numberand length of piles. These results may be used to assess the effects of group interaction inrelation to pile spacing. It should be noted that the model piles were not instrumented todetermine the effects of interaction on shaft and base capacity separately and the observedgroup reduction factors have been defined in terms of overall capacity.

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The contribution of a ground-bearing cap to the group capacity may be calculated usingthe approximate method given in Section 6.3.2(3).

6.3.4 Vertical Pile Group in Rock under Compression

The overall capacity of a pile group founded on rock or a group of rock sockets canbe taken as the sum of the individual pile capacities (i.e. with a group efficiency factor ofunity).

6.3.5 Vertical Pile Group under Lateral Loading

For a laterally-loaded group of vertical piles, similar checks for the sum of individualpile lateral capacities and for block or row failure should be made as for vertical loading.

Prakash (1962) found from model tests in sand that piles behave as individual units ifthe centre-to-centre spacing is more than three pile widths in a direction normal to the lineof the loading and where they are spaced at more than six to eight pile widths measured alongthe loading direction. These findings are supported by results of finite element analysesreported by Yegian & Wright (1973) who show that, for a given pile spacing, the groupreduction factor of a row of piles is smaller (i.e. greater interaction) when the horizontalloading is applied along the line joining the piles, compared to that when the loading isperpendicular to the line joining the piles.

Poulos & Davis (1980) summarised the results of model tests carried out on pilegroups in sand and clay soils respectively. These indicate a group reduction factor for lateralloading of about 0.4 to 0.7 for a spacing to diameter ratio of between 2 and 6. Results ofinstrumented full scale tests on a pile group in sand reported by Brown et al (1988) indicatethat the lateral capacity of piles in the leading row is about 90% of that of a single pile;however, the measured capacity of the piles in the trailing row is only about 40% of a singlepile. This is attributed to the effects of 'shadowing1, i.e. effects of interaction of stress fieldsin the direction of the load (see also discussion in Section 6.6.2(3)).

The effect of possible interaction of piles constructed by different techniques in a groupon the lateral capacity of a pile group has not been studied systematically.

Both Elson (1984) and Fleming et al (1992) suggested that a pragmatic approach maybe adopted and recommended that the group reduction factor may be taken as unity wherethe centre-to-centre pile spacing is equal to or greater than three pile diameters alongdirections parallel and perpendicular to the loading direction. For a group of closely-spacedpiles (spacing/diameter less than 3), the group may be considered as an equivalent single pile.

There are clearly differing views in the literature on the group reduction factor for alaterally-loaded pile group. In practice, it is the group lateral deflection or the structuralcapacity of the pile section that governs the design, with the possible exception of short rigidpiles. It is therefore considered that the recommendations by Fleming et al (1992) canreasonably be adopted for practical purposes, except for short rigid piles (see Figure 15 forcriteria for short rigid piles), where reference may be made to the findings by Poulos &

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Davis (1980) described above.

In evaluating the block or row failure mechanism, both the side shear and the baseshear resistance should be considered.

For rock-socketed piles, possible joint-controlled failure mode should be consideredand a detailed assessment of the joint pattern must be made.

The bending moment and shear force induced in the piles should be checked to ensurethat the ultimate resistance is not governed by the structural capacity. For routine design ofpile groups with piles having similar bending stiffness, the simplifying assumption that eachpile will carry an equal share of the applied horizontal load may be made. Where the pilestiffnesses vary significantly, a detailed frame analysis may be carried out to assess the forcedistributions.

6.3.6 Vertical Pile Group under Tension Loading

The uplift capacity of a pile group is the lesser of the following two values :

(a) the sum of uplift resistance of individual piles withallowance for interaction effects, and

(b) the sum of the shear resistance mobilised on the surfaceperimeter area of the group and the effective weight ofsoil/piles enclosed by this perimeter.

In assessing the block failure mechanism, the group effect could reduce the verticaleffective stress in the soil and the influence of this on the shaft resistance may need to beconsidered.

For driven piles in granular soils, densification effects as discussed in Section 6.3.2will be relevant. It is considered that the group reduction factor in this case may be assumedto be unity. For bored piles in granular soils, the results of model tests carried out byMeyerhof & Adams (1968) as summarised in Figure 34 may be used to help assess theappropriate group reduction factor.

For piles in clays, results of model tests carried out by Meyerhof & Adams (1968)indicate that the group reduction factors for uplift are in reasonable agreement with thosereported by Whitaker (1957) for piles under compression. The results shown in Figure 33may therefore be used for pile groups in clays under tension.

6.3.7 Pile Groups Subject to Eccentric Loading

Where the applied load is eccentric, there is a tendency for the group to rotate whichwill be resisted by an increase in horizontal soil pressures. However, when the passive soilpressure limits are reached, a substantial reduction in the group capacity could occur.

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Broms (1981) suggested an approximate method for determining the ultimate capacityof a general pile group, which comprises a combination of vertical and raking piles, whenit is subject to an eccentric vertical load. This formulation reduces the problem to a staticallydeterminate system and is a gross simplification of the interaction problem. The applicabilityof this proposed methodology is uncertain and is not proven.

Early model tests were carried out by Meyerhof (1963) for pile groups in clays andby Kishida & Meyerhof (1965) for pile groups in granular soils. These were supplementedby more recent model tests reported by Meyerhof & Purkayastha (1985) on the ultimatecapacity of pile groups under eccentric vertical loading and inclined loading. These testswere carried out in a layered soil consisting of clay of varying thicknesses over sand. Theresults were expressed as polar group efficiency diagrams for different ratios of clay to sandthickness. In the absence of field data, the test results summarised in Figure 35 may be usedas a basis for making an approximate allowance for the reduction in ultimate capacity of apile group subjected to eccentric and/or inclined loading.

Alternatively, the load and capacity of individual piles may be considered. Asimplified and commonly-used method for determining the distribution of loads in individualpiles in a group subject to eccentric loading is the 'rivet group1 approach (Figure 36). Itshould be noted that the load distribution in the piles determined using this method may notbe a good representation of the actual distribution in the group due to interaction effects,particularly where there are raking piles.

In assessing the effects of pile-soil-pile interaction on individual pile capacities, theguidance given in Sections 6.3.2 to 6.3.6 for group reduction factors for vertical pile groupssubject to axial loads and lateral loads respectively may also be taken to apply to general pilegroups for practical purposes.

When a pile group is subject to an eccentric horizontal load, torsional stresses incombination with bending stresses will be transmitted to the piles. The behaviour of aneccentrically-loaded pile group is poorly understood. Where there is a pile cap, a proportionof the load effect will be supported by mobilisation of passive pressure on the cap withoutbeing transferred to the piles. Reference may be made to Randolph (1981b) for analysis ofpile behaviour under torsion loading.

6.4 NEGATIVE SKIN FRICTION ON PILE GROUPS

As far as negative skin friction is concerned, group interaction effects are beneficialin that the downdrag acting on individual piles will be reduced. The possible exception is forsmall pile groups (say less than five piles) in very soft soils undergoing substantial settlementsuch that slip occurs in all the piles, resulting in no reduction in downdrag compared to thatof a single pile. It should be noted that the distribution of downdrag forces between piles willnot be uniform, with the centre piles experiencing the least negative skin friction due tointeraction effects.

For practical purposes, the limiting downdrag may be taken as the lesser of :

(a) the sum of negative skin friction around pile group

perimeter and effective weight of ground enclosed by theperimeter, and

(b) the sum of negative skin friction on individual piles (with acautious allowance for interaction effects).

Wong (1981) reviewed the various analytical methods and put forward an approachbased on the assumption that the settling soil is in a state of plastic failure as defined by theMohr-Coulomb criterion. In this method, allowance can be made for group action, effect ofpile spacing and arching on the vertical effective stress, together with the different stresscondition for piles at different positions in a group.

For an internal pile (i.e. piles not along the perimeter of the group), the negative skinfriction will be limited to the submerged weight of the soil column above the neutral point(Section 5.8.2) as this is the driving force.

Kuwabara & Poulos (1989) carried out a parametric study on the magnitude anddistribution of downdrag using the boundary element method. It was shown that the methodgave reasonable agreement with observed behaviour for a published field experiment in Japan.

The above methods are capable of predicting the distribution of negative skin frictionin a large pile group and hence assess the average downdrag on the group. For pile groupsof five piles or more at a typical spacing of three to five pile diameters, interaction effectswill result in a reduction in the average downdrag. Analysis using the above methodstogether with available overseas instrumented full-scale data (e.g. Okabe, 1977; Inoue, 1979)indicate that the reduction can be in the range of 15% to 30%.

In the absence of instrumented data in Hong Kong, it is recommended that a generalreduction of 10% to 20% on the negative skin friction in a single pile within a group may beconservatively assumed for design purposes, for a pile group consisting of at least five pilesat customary spacing. The appropriate value to be adopted will depend on the spacing andnumber of piles in a group.

Where the calculated reduction in negative skin friction due to group effects is inexcess of that observed in field monitoring, consideration should be given to making a morecautious allowance or instrumenting the piles in order to verify the design assumptions.

The effect of negative skin friction may lead to reduction in the effective overburdenpressure and hence the capacity of the bearing stratum. Davies & Chan (1981) developed ananalysis put forward by Zeevaert (1959) which makes allowance for the reduction in effectiveoverburden pressure acting on the bearing stratum as a result of arching between piles withina pile group.

6.5 DEFORMATION OF PILE GROUPS

6.5.1 Axial Loading on Vertical Pile Group

(1) General Based on linear elastic assumptions, the ratio of immediate settlement

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to total settlement of a pile group is expected to be less than that for a single pile. Generally,the ratio is in the range of 2/3 to 3/4 for typical friction-pile group configurations in granularsoils (Poulos & Davis, 1980). For end-bearing groups, the relative amount of immediatesettlement is generally greater than for friction pile groups. Pile interaction generally resultsin a higher percentage of the total load being transferred to the base of piles compared to thatin isolated piles.

The settlement of a pile group subject to a given average load per pile is generallylarger than that in a single pile under the same load. The corresponding ratio is termed thegroup settlement ratio (Rgs). Group settlement ratios observed in full scale tests on pilegroups founded in granular soils are summarised by O'Neill (1983). It was found that Rgs

is generally larger than unity, except where driven piles have been installed into loose sand,increasing the ground stiffness due to densification effects.

The guidance given in Section 5.13.2 on soil stiffness also applies to settlementpredictions for a pile group. The stress bulb associated with a pile group will be larger thanthat for a single pile and the settlement characteristics will therefore be influenced by soilsat greater depths.

The various approaches which have been proposed for assessing pile group settlementmay be categorised as follows :

(a) semi-empirical methods,

(b) equivalent raft (or pier) method,

(c) interaction factor methods, and

(d) numerical methods.

The analysis of the settlement of a pile group incorporating a ground-bearing cap isdiscussed in Section 6.6.3.

(2) Semi-empirical methods. Various semi-empirical formulae derived from limitedfield observations (e.g. Skempton, 1953; Vesic, 1969; Meyerhof, 1976) have been proposedfor predicting settlement of pile groups in sand. A commonly-used rale-of-thumb is toassume the differential settlement of the pile group is up to half the maximum groupsettlement in uniform soils.

The empirical formulae suffer from the drawback that they have not been calibratedagainst observations made in Hong Kong and their formulation lacks a sound theoreticalbasis, and therefore their use is not recommended for detailed design.

(3) Equivalent raft (orpier) method. The equivalent raft (or pier) method is a widely-used simplified technique for the calculation of pile group settlement. In this method, the pilegroup is idealised as an equivalent raft which is assumed to be fully flexible. The locationand size of the equivalent raft is dependent on the mode of load transfer, i.e. whether theapplied load is resisted primarily in shaft friction or end bearing (Figure 37). A more recentdevelopment of the equivalent raft concept is discussed by Randolph (1994).

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The settlement of the equivalent raft can be calculated using elasticity solution forgranular soils and consolidation theory for clays. The settlement at pile top is obtained bysumming the raft settlement and the elastic compression of the pile length above theequivalent raft. An assessment may be made of the influence of the relative rigidity of a rafton settlement following Fraser & Wardle (1976). Depth and rigidity corrections factors maybe applied to the calculated settlement as appropriate (Tomlinson, 1994; Davis & Poulos,1968).

This approach is generally adequate for routine calculations involving simple pilegroup geometries to obtain a first order estimate of group settlement. However, it does notconsider the influence of pile spacing or effect of pile interaction in a rational manner. Also,the effects of relative stiffness between the structure and foundation are accounted for in onlyan approximate manner with the use of a rigidity correction factor. Thus, the method shouldbe used with caution for the analysis of pile groups with a complex geometry, greatlydifferent pile lengths, or where the loading is highly non-uniform.

(4) Interaction factor methods. A widely used method of analysing the pile groupsettlement is based on the concept of interaction factors ($) defined as follows :

, _ additional settlement caused by an adjacent pile under load (f.

settlement of pile under its own load

This is an extension of the elastic continuum method for analysis of settlement ofsingle piles where the interaction effects in a pile group are assessed by superposition. Basicsolutions for the group settlement ratio (Rgs) for incompressible friction or end-bearing pilegroups are summarised by Poulos & Davis (1980). Correction factors can then be appliedfor base enlargement, depth to incompressible stratum, non-homogeneous soil, effect of pileslip, interaction between piles of different sizes, pile compressibility, rigidity of the bearingstratum, etc.

An alternative and simplified form of the interaction factor method was proposed byRandolph & Wroth (1979). Equations have been derived for shaft and base interaction factorrespectively as summarised in Figure 38. The relationship between group settlement ratio,Rgs and the number of piles is shown for two simple cases in Figures 39(a) & (b). It can beseen that interaction effects are less pronounced in a soil with increasing stiffness with depththan in a homogeneous soil.

The assumption of linear elasticity for soil behaviour is known to over-estimateinteraction effects in a pile group. Jardine et al (1986) demonstrated the importance of non-linearity in pile group settlement and load distribution with the use of finite element analyses.

Poulos (1988b) has modified the interaction factor method to incorporate the effectsof strain-dependency of soil stiffness. The modified analysis shows that the presence ofstiffer soils between piles results in a smaller group settlement ratio and a more uniform loaddistribution than that predicted based on the assumption of a linear elastic, laterallyhomogeneous soil.

It should be noted that in the formulation of the interaction factor method, thereinforcing effect of the piles on the soil mass is disregarded. This assumption becomes less

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realistic for sizeable groups of piles with a large pile stiffness factor K (equation 5.12).

Fraser & Lai (1982) reported comparisons between the predicted and monitoredsettlement of a group of driven piles founded in granitic saprolites. The prediction was basedon the elastic continuum method which was found to over-estimate the group settlement byup to about 100% at working load even though the prediction for single piles comparesfavourably with results of static load tests. Similar findings were reported by Leung (1988).This may be related to the densification effect associated with the installation of driven pilesor the over-estimation in the calculated interaction effect by assuming a linear elastic soil.

In general, the interaction factor method based on linear elastic assumptions should,in principle, give a conservative estimate of the magnitude of the pile group settlement. Thisis because the interaction effects are likely to be less than assumed.

(5) Numerical methods. A number of approaches based on numerical methods havebeen suggested for a detailed assessment of pile group interaction effects. Although these areunlikely to be of direct use for routine design problems, they may nonetheless provide auseful insight into the mechanism of behaviour. The designers should be aware of thecapability and limitations of the available methods where their use is considered justifiablefor complex problems. Examples of where numerical methods can be applied more readilyin practice include design charts based on these methods for simple cases, which may berelevant for the design problem in hand. Some such design charts are discussed in thefollowing, together with the common numerical methods that have been developed forfoundation analysis.

A more general solution to the interaction problem was developed by Butterfield &Bannerjee (1971a) using the boundary element method. Results generally compare favourablywith those derived using the interaction factor method (Hooper, 1979). An alternativeapproach is to replace the pile group by a block of reinforced soil in a finite element analysis(Hooper & Wood, 1977).

Butterfield & Douglas (1981) summarised the results of boundary element analyses ina collection of design charts. The results are related to a stiffness efficiency factor (Rg)which is defined as the ratio of the overall stiffness of a pile group to the sum of individualpile stiffness. This factor is equal to the inverse of the group settlement ratio (i.e. Rg =1/Rgs). Fleming et al (1992) noted that the stiffness efficiency factor is approximatelyproportional to the number of. piles, plotted on a logarithmic scale, i.e. Rg = np"

a where np

is the number of piles. Typical design charts for calculating the value of the exponent a aregiven in Figure 40. For practical problems, the value of a usually lies in the range of 0.4to 0.6. It is recommended that this simplified approach may be used for pile groups withsimple geometry, i.e. regular arrangement of piles in a uniform soil.

Recent developments in numerical methods include the infinite layer method forlayered soils (Cheung et al, 1988) and the formulation proposed by Chow (1989) for cross-anisotropic soils. Chow (1987) also put forward an iterative method based on a hybridformulation which combines the load transfer method (Section 5.13.2(2)) and elasticcontinuum approach (Section 5.13.2(3)) for single piles using Mindlin's solution to allow forgroup interaction effects.

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Results of numerical analysis of the settlement of a pile group that are socketed intoa bearing stratum of finite stiffness are presented by Chow et al (1990) in the form of designcharts.

Computer programs based on the 'beam (or slab) on spring foundation' model may beused where springs are used to model the piles and the soil (Sayer & Leung, 1987; Stubbings& Ma, 1988). This approach can reasonably be used for approximate foundation-structureinteraction analysis. For a more detailed and rational assessment of the foundation-structureinteraction and pile-soil-pile interaction, iterations will be necessary to obtain the correct non-uniform distribution of spring stiffness across the foundation to obtain compatible overallsettlement profile and load distribution between the piles (Figure 41).

There is a relatively wide range of approaches developed for detailed studies ofinteraction effects on the settlement of a pile group. Different formulations are used and itis difficult to have a direct comparison of the various methods. The applicability andlimitations of the methods for a particular design problem should be carefully considered andthe chosen numerical method should preferably be calibrated against relevant case historiesor back analysis of instrumented behaviour. In cases where a relatively unfamiliar orsophisticated method is used, it would be advisable to check the results are of a similarmagnitude using an independent method.

6.5.2 Lateral Loading on Vertical Pile Group

(1) General The assessment of the lateral deflection of a pile group is a difficultproblem. The response of a pile group involves both the lateral load-deformation and axialload-deformation characteristics as a result of the tendency of the group to rotate when loadedlaterally. Only when the rotation of the pile cap is prevented would the piles deflect purelyhorizontally.

As a general guideline, it may be assumed that piles can sustain horizontal loads of upto 10% of the allowable vertical load without special analysis (Canadian Geotechnical Society,1992) unless the soils within the upper 10% of the critical length of the piles (see Sections5.13.3(3) & 5.13.3(4) for discussion on critical length) are very weak and compressible.

(2) Methodologies for analysis. There are proposals in the literature for empiricalreduction factors for % (Table 15) to allow for group effects in the calculation of deflection,shear force, bending moment, etc. using the subgrade reaction method. Although thesesimplifying approximations do not have a rational theoretical basis in representing the highlyinteractive nature of the problem, in practice they are generally adequate for routine designproblems and form a reasonable basis for assessing whether more refined analysis iswarranted. .

An alternative approach which may be used for routine problems is the elasticcontinuum method based on the concept of interaction factors as for the calculation of pilegroup settlement. Elastic solutions for a pile group subject to horizontal loading aresummarised by Poulos & Davis (1980).

Based on the assumptions of a linear elastic soil, Randolph (1981a) derived expressions

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for the interaction factors for free-headed and fixed-headed piles loaded laterally (Figure 42).It can be deduced from this formulation that the interaction of piles normal to the applied loadis only about half of that for piles along the direction of the load. The ratio of the averageflexibility of a pile group to that of a single pile for lateral deflection under the condition ofzero rotation at ground level can also be calculated. This ratio, defined as the group lateraldeflection ratio (Rh), is analogous to the group settlement ratio (Rgs). As an illustration,results for typical pile group configurations are shown in Figure 39 which illustrates that thedegree of interaction under lateral loading is generally less pronounced compared to that forvertical loading. This approach by Randolph (198la) is simple to use and is consideredadequate for routine problems where the group geometry is relatively straight forward.

A further alternative is to carry out an elasto-plastic load transfer analysis using thesubgrade reaction method with an equivalent pile representing the pile group. In thisapproach, the group effect can be allowed for approximately by reducing the soil resistanceat a given deflection or increasing the deflection at a given soil pressure (Table 15). Inpractice, the actual behaviour will be complex as the effective H~6h curve for individual pilesmay be different and dependent on their relative positions in the pile group. Considerablejudgement is required in arriving at the appropriate model for the analysis for a givenproblem.

(3) Effect of pile cap. Where there is a pile cap, the applied horizontal loads will beshared between the cap and the pile as a function of the relative stiffness. The unitdisplacement of the pile cap can be determined following the solution given by Poulos &Davis (1974), whereas the unit displacement of the piles may be determined using themethods given in Sections 5.13.3 and 6.5.2(2). From compatibility considerations, the totaldisplacement of the system at pile head level can be calculated and the load split between thecap and the piles determined. Care should be taken to make allowance for possible yieldingof the soil where the strength is fully mobilised, after which any additional loading will haveto be transferred to other parts of the system.

Kim et al (1977) observed from full scale tests on a group of vertical piles that theeffect of contact between a ground-bearing cap and the soil is to reduce the group deflectionby a factor of about two at working conditions. However, it was reported by O'Neill (1983)that the effect of cap contact is found to be negligible where the majority of the piles arebattered.

6.5.3 Combined Loading on General Pile Group

(1) General Deformations and forces induced in a general pile group comprisingvertical and raking piles under combined loading condition are not amenable to presentationin graphical or equation format. A detailed analysis will invariably require the use of acomputer.

(2) Methodologies for ana/^/5. Historically, simple groups of piles have beenanalysed by assuming that the piles act as structural members. In this method, either a directresolution of forces is made where possible or a structural frame analysis is carried out(Hooper, 1979). The presence of the soil is accounted for approximately by assuming aneffective pile length. This is a gross simplification of the complex relative stiffness problem

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in a soil continuum and should be used with extreme caution.

A more rational approach is to model the soil as an elastic continuum. A number ofcommercial computer programs have been written for general pile group analysis based onidealising the soil as a linear elastic material, e.g. PIGLET (Randolph, 1980), DEFPIG(Poulos, 1990a), PGROUP (Bannerjee & Driscoll, 1978), etc. which have been applied toproblems in Hong Kong. The first two programs are based on the interaction factor methodwhile the last one uses the boundary element method. A brief summary of the features ofsome of the computer programs developed for analysis of general pile groups can be foundin Poulos (1989a) and the report by the Institution of Structural Engineers (1989). Computeranalyses based on the elastic continuum method generally allow more realistic boundaryconditions, variation in pile stiffness and complex combined loading to be modelled.

Comparisons between results of different computer programs for simple problems havebeen carried out, e.g. O'Neill & Ha (1982) and Poulos & Randolph (1983). The comparisonsare generally favourable with discrepancies which are likely to be less than the margin ofuncertainty associated with the input parameters. Comparisons of this kind lend confidencein the use of these programs for more complex problems.

Pile group analysis programs can be useful to give an insight into the effects ofinteraction and to provide a sound basis for rational design decisions. In practice, however,the simplification of the elastic analyses, together with the assumptions made for theidealisation of the soil profile, soil properties and construction sequence could potentially leadto misleading results for a complex problem. Therefore, considerable care must be exercisedin the interpretation of the results.

The limitations of the computer programs must be understood and the idealisations andassumptions made in the analyses must be compatible with the problem being considered.It would be prudent to carry out parametric studies to investigate the sensitivity of thegoverning parameters for complex problems.

(3) Choice of parameters. One of the biggest problems faced by a designer is thechoice of appropriate soil parameters for analysis. Given the differing assumptions andproblem formulation between computer programs, somewhat different soil parameters maybe required for different programs for a certain problem. The appropriate soil parametersshould ideally be calibrated against a similar case history or derived from the back analysisof a site-specific instrumented pile test using the proposed computer program for a detailedanalysis.

6.6 DESIGN CONSIDERATIONS IN SOIL-STRUCTURE INTERACTIONPROBLEMS

6.6.1 General

In practice, piles are coupled to the structure and do not behave in isolation. Soil-structure interaction arises from pile-soil-pile interaction and pile-soil-structure interaction.The interaction is a result of the differing stiffnesses which govern the overall load-deformation characteristics of the system as movements and internal loads re-adjust under the

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applied load.

Interaction also occurs in situations where piles are installed in a soil undergoingmovements. The presence of stiff elements (i.e. the piles) will modify the free-field groundmovement profile which in turn will induce movements and forces in the piles.

The proper analysis of a soil-structure interaction problem is complex and generallyrequires the use of a computer which must incorporate a realistic model for the constitutivebehaviour of the soil. The computational sophistication must be viewed in perspective of theapplicability of the simplifying assumptions made in the analysis and the effects of inherentheterogeneity of the ground, particularly for soils and rocks in Hong Kong. The results ofthe analyses should be used as an aid to judgement rather than as the sole basis for designdecisions.

In practice, it is unusual to carry out detailed soil-structure interaction analyses forroutine problems. However, a rational analytical framework is available (e.g. elasto-plasticfinite element analysis) and could be considered where time and resources permit and forcritical or complex design situations. In addition, the analysis could be used for backcalculation of monitored behaviour to derive soil parameters.

6.6.2 Load Distribution between Piles

(1) General A knowledge of the load distribution in a pile group is necessary inassessing the profile of movement and the forces in the pile cap. Linear elastic methods areusually used for this purpose although the predictions tend to over-estimate the loaddifferentials.

(2) Piles subject to vertical loading. The distribution of vertical loads in a free-standing pile group with a rigid pile cap is predicted to be non-uniform by continuumanalyses assuming a linear elastic soil (Poulos & Davis, 1980). Piles near the centre of agroup are expected to carry less loads than those at the edges. It is however incorrect todesign for this load re-distribution by increasing the capacity of the outer piles in order tohave the same factor of safety as for a pile loaded singly. This is because the stiffness of theouter piles would then increase, thereby attracting more load.

The general predicted pattern of load distribution has been confirmed by measurementsin model tests and field monitoring of prototype structures for piles founded in clayey soils.Typically, the measurements suggest that the outer piles could carry a load which is aboutthree to four times that of the central piles at working load conditions in a large pile group(Whitaker,1957; Sowers et al, 1961; Cooke, 1986).

For groups of displacement piles in granular soils, a different pattern was reported.Measurements made by Vesic (1969) in model tests involving jacked piles indicate a differentload distribution to that predicted by elastic theory, with the centre piles carrying between20% and 50% more load than the average load per pile. The distribution of the shaft frictioncomponent is however more compatible with elastic continuum predictions (i.e. outer pilescarrying the most load). The effects of residual stresses and proximity of the boundaries ofthe test chambers on the results of these model tests are uncertain (Kraft, 1991). Beredugo

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(1966) and Kishida (1967) also studied the influence of the order of installing driven piles andfound that, at working conditions, piles that have been installed earlier tend to carry less loadthan those installed subsequently.

At typical working loads, the load distribution for a pile group in granular soils islikely to be similar to that in clays, particularly for bored piles. This is supportedqualitatively by results of model tests on instrumented strip footings bearing on sand reportedby Delpak et al (1992). Their model test results indicate that at working load conditions thedistribution of contact pressure is broadly consistent with elastic solutions, whereas at thecondition approaching failure the central portion shows the highest contact pressure.

The non-uniform load distribution can be important where the mode of pile failure isbrittle, e.g. for piles end-bearing in granular soils overlying a weaker layer where there isa risk of punching failure. The possibility of crushing or structural failure of the pile shaftshould also be checked for piles, particularly for mini-piles.

(3) Piles subject to lateral loading. For piles subject to lateral loading, centrifugetests on model pile groups in sand showed that the leading piles carried a slightly higherproportion of the overall applied load than the trailing piles (Barton, 1982). The load splitwas of the order of 60%/40% at working conditions. Similar findings were reported bySelby & Poulos (1984) who concluded that elastic methods are not capable of reproducing theresults observed in model tests.

Ochoa & O'Neill (1989) observed from full-scale tests in sand that 'shadowing' effects(i.e. geometric effects that influence the lateral response of individual piles), together withpossible effects due to the induced overturning moment, can significantly affect thedistribution of forces in the piles. Both the soil resistance and the stiffness of a pile in atrailing row are less than those for a pile in the front row because of the presence of the pilesahead of it. These effects are not modelled in conventional analytical methods, i.e. elasticcontinuum or subgrade reaction methods. Nevertheless, it was found that the elasticcontinuum method gave reasonable predictions of the overall group deflection, although notso good for predictions of load and moment distribution for structural design under workingconditions. An empirically-based guideline is given by the New Zealand Ministry of Worksand Development (1981) for the reduction in the modulus of horizontal subgrade reaction (Kh)for the trailing piles where the pile spacing is less than eight pile diameters along the loadingdirection.

Brown et al (1988) found from instrumented field tests that the applied load wasdistributed in greater proportion to the front row than to the trailing row by a factor of abouttwo at maximum test load but the ratio is less at smaller loads. This resulted in largerbending moment in the leading piles at a given loading.

In contrast, results of model pile tests in clay indicate an essentially uniform sharingof the.applied load between the piles (Fleming etal, 1992). Brown et al (1988) also foundthat the 'shadowing' effect is much less significant in the case of piles in clay than in sand.

The actual distribution of loads between piles at working condition is dependent on thepile group geometry and the relative stiffness between the cap, the piles and the soil. Thisis important in evaluating the deflection profile and structural forces in the cap and the

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superstructure.

For design purposes, the assumption that the applied working load is shared equallyby the piles may be made for a uniform pile group. Where the pile group consists of pilesof different dimensions, the applied lateral load should be distributed in proportion to thestiffness as follows :

(6.3)

where Hxi = horizontal load on pile iHx = total horizontal load in x-directionIyi = moment of inertia of ith pile about its y-axis

In general, as long as the pile length is larger than the critical pile length under lateralloading for a given soil (Section 5.13.3(4)), the group behaviour under lateral loading of agroup of piles of differing lengths will not be different from a group of piles of equal lengths.

6.6.3 Behaviour of Piled Rafts at Working Condition

(1) General. Where a rigid piled raft is resting on competent ground, a proportionof the applied vertical load will be transferred to the soil by bearing. This will affect the loaddistribution in the piles and the overall settlement behaviour of the foundation. The raftshould be adequately designed against punching failure.

Field measurements of the proportion of load taken by the raft and the piles at workingconditions are summarised by Hooper (1979) and Cooke (1986). These suggest that the ratioof load in the most heavily loaded piles in the perimeter of the group to that in the leastheavily loaded pile near the centre could be about 2.5.

It is likely that the load distribution in a group of bored piles founded in granular soilsis similar to that in clayey soils at working load.

(2) Methodologies for analysis. Some of the approaches which have been proposedto analyse the behaviour of a piled raft are summarised by Hooper (1979) and in the reportby the Institution of Structural Engineers (1989).

The interaction factor approach discussed in Section 6.5.1(4) can be used (Poulos &Davis, 1980). For most practical problems, the influence of pile cap contact on the overallfoundation stiffness is not significant at working condition.

Fleming et a l( 1992) presented a. simple method of analysing the combined stiffnessof the raft and the piles which allows for interaction between the piles and the raft(Figure 43). The effect of alternative piling layout on foundation settlement can be assessed.

It is common practice in Hong Kong for foundations incorporating piles and a ground-bearing raft to be designed assuming the applied structural load is taken entirely by the piles.Where the raft is bearing on competent ground, this assumption will be conservative. In this

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case, it is suggested that a more realistic analysis may be carried out using the methodsuggested by Fleming et al (1992).

Numerical techniques for detailed analyses of piled raft foundations are described byHongladaromp et al (1973) and Hain & Lee (1978). These analyses are complex and notwarranted for most practical problems.

(3) Field behaviour. Leung & Radhakrishnan (1985) reported the behaviour of aninstrumented piled raft founded on weathered sedimentary rock in Singapore. The loaddistribution between the raft and the piles was found to be about 60% 740% at the end ofconstruction. The measured raft pressures were highest below the centre of the raft.However, the degree of non-uniformity of the applied load is not known.

Radhakrishnan & Leung (1989) reported, for a raft supported on rock-socketed piles,that the load transfer behaviour during construction differed from the behaviour during theload test, with less skin friction mobilised over the upper three diameters of the pile shaftunder construction load. It was postulated by Radhakrishnan & Leung (1989) that thepresence of the rigid pile cap may have inhibited the development of friction over the upperpile shaft. The base resistance mobilised under long-term structural loads was also noted tobe significantly higher than that under the pile test. This may be due to group interactioneffects or creep of the concrete. To a certain extent, the behaviour will also be affected bythe ground conditions of the test pile site.

6.6.4 Use of Piles to Control Foundation Stiffness

A recent development in the field of soil-structure interaction is the concept of the useof optimal pile configuration to control the overall foundation stiffness in order to minimisedifferential settlement and variations in the structural forces.

The concept was developed for piled rafts and is based on controlling the re-distribution of load through the introduction of a limited number of piles positionedjudiciously. The concept can be applied to cases where the raft bears on a competent stratumand the piles are only required for controlling settlements, not for overall bearing capacity.In this case, the resistance of the piles can be designed to be fully mobilised at workingcondition, thus taking a proportion of the applied load away from the raft. Piles may alsobe positioned below concentrated loads in order to minimise the bending of the raft by takinga share of the applied load. In principle, the concept also works for a free-standing pilegroup with a rigid cap where piles can be positioned judiciously such that a more uniformload distribution and hence settlement profile is achieved. Experimental studies of thebehaviour of piled rafts are described by Long (1993).

Burland & Kalra (1986) described a successful field application of this concept butwarned that the approach should be considered only for friction piles in clays and not forpiles bearing on a strong stratum such as rock or gravel where the mode of failure could bebrittle and uncontrolled. Simpson et al (1987) further warned that the use of these'settlement-reducer' type piles may give rise to problems of large local differential movementsin the case of a general rise in the groundwater table.

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The concept of using piles to manipulate the overall foundation stiffness has also beenapplied to the design of approach embankments for bridges. In this case, piles with smallcaps are similarly designed to have their resistance fully mobilised. These piles are referredto as the BASF (Bridge Approach Support Piling) system by Reid & Buchanan (1983) andare used in conjunction with a continuous geotextile mattress over the tops of the pile capsin order to reduce the embankment settlement.

Hewlett & Randolph (1988) developed a method of analysis for piled embankmentsbased on assumed arching mechanisms. This method can be used to optimise the number ofpiles required to reduce the settlement of an embankment.

In general, the concept of using piles to control foundation stiffness requires anaccurate assessment of the distribution of pile loads and settlement profile. In view of thehighly heterogeneous nature of the corestone-bearing weathering profiles in Hong Kong, suchconcepts should be applied with caution. The validity of the approach will need to be verifiedby means of sufficient load tests and monitoring of prototype structures.

6.6.5 Piles in Soils Undergoing Movement

(1) General Loads can be induced in piles installed in a soil which undergoesdeformation after pile construction. A common situation arises where bridge abutment pilesinteract with the soft soil which deforms both vertically and laterally as a result ofembankment construction. The use of raking piles in such situations should be avoided asthere is a risk of the structural integrity of the piles being impaired due to excessive groundsettlements. Stabilising piles are sometimes used to enhance the factor of safety ofmarginally-stable slopes (Powell et al, 1990) and forces will be mobilised in these piles whenthere is a tendency for the ground to move.

This class of interaction problem is complicated and the behaviour will, in part, bedependent on the construction sequence of the piles and the embankment, pile groupgeometry, consolidation behaviour, free-field deformation profile, relative stiffness of thepile and the soil, etc,

(2) Piles in soils undergoing lateral movement. For the problem of bridge abutmentpiles, Hambly (1976) discussed various methods of analysis and cautioned against the use ofsimple elastic continuum methods for problems involving large deformation.

Poulos & Davis (1980) proposed a simplified elastic approach based on interaction ofthe moving soil and the piles with allowance made for the limiting pressure that the soil mayexert on the pile. The use of this method requires an estimate of the free field horizontal soilmovement profile. NAVFAC (1982) suggested a simplified hand method of calculating thedistribution of pressure along 'stabilizing' piles based on the work reported by De Beer &Wallays (1972). These methods can be used for .conceptual designs.

Based on observations made in centrifuge tests, simple design charts have been putforward by Springman & Bolton (1990) for assessing the effect of asymmetrical surchargeloading adjacent to piles. It is suggested that this approach can be used for routine designproblems in so far as they are covered by the charts.

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Stewart et al (1992) reviewed a range of available simplified design methods andconcluded that they are generally inconsistent although some aspects of the observedbehaviour can be accounted for to a varying degree by the different methods. For complexproblems, a more sophisticated numerical analysis (e.g. finite element method) may benecessary.

(3) Piles in heaving soils. Tension forces will be developed in piles if the soil heavessubsequent to pile installation (e.g. piles in a basement prior to application of sufficientstructural load). The simplified method of analysis presented by O'Reilly & Al-Tabbaa(1990) may be used for routine design. The analysis can also take into account progressivecracking in pile with increase in loading by making allowance for possible reduction in pilestiffness (and hence reduction in pile tension).

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7. PILE INSTALLATION AND CONSTRUCTION CONTROL

7.1 GENERAL

There are uncertainties in the design of piles due to the inherent variability of theground conditions and the potential effects of the construction process on pile performance.

Adequate supervision must be provided to ensure the agreed construction method isfollowed and enable an assessment of the actual ground conditions to be carried out duringconstruction. It is necessary to verify that the design assumptions are reasonable.

Foundation construction is usually on the critical path and the costs and time delayassociated with investigating and rectifying defective piles could be considerable. It istherefore essential that pile construction is closely supervised by suitably qualified andexperienced personnel who fully understand the assumptions on which the design is based.Detailed construction records must be kept as these can be used to identify potential defectsand diagnose problems in the works.

This chapter summarises the equipment used in the construction of the various typesof piles commonly used in Hong Kong. Potential problems associated with the constructionof piles are outlined and good construction practice is highlighted. The range of controlmeasures and available engineering tools, including integrity testing, that could be used tomitigate construction problems and identify anomalies in piles are presented. It should benoted that the range of problems discussed is not exhaustive. It is important that thedesigners should carefully consider what could go wrong and develop a contingency planwhich should be reviewed regularly in the light of observations of the works as they proceed.

7.2 INSTALLATION OF DISPLACEMENT PILES

7.2.1 Equipment

Displacement piles are installed by means of a driving hammer or a vibratory driver.There are a range of hammer types including drop hammer, steam or air hammer, dieselhammer and hydraulic hammer. Use of these hammer types are classified as percussivepiling, which is subject to the requirements of Noise Control Ordinance (Hong KongGovernment, 1994).

It is important to exercise directional control and maintain the pile in alignment duringinitial pitching and driving. Leaders held in position by a crane are suitable for support ofboth the pile and the hammer during driving, and may be used for vertical and raking piles.Alternatively, vertical piles may be supported in a trestle or staging and driven with ahammer fitted with guides and suspended from a crane.

Where a hammer is used to produce impacts on a precast concrete pile, the headshould be protected by an assembly of dolly, helmet and packing or pile cushion (Figure 44).The purpose of the assembly is to cushion the pile from the hammer blows and distribute thedynamic stresses evenly without allowing excessive lateral movements during driving. In

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addition, the life, of the hammer would be prolonged by reducing the impact stresses.Packing (or pile cushion) is generally not necessary for driving steel piles.

A follower is used to assist driving in situations where the top of the pile is out ofreach of the working level of the hammer. Wong et al (1987) showed that where theimpedance of the follower matches that of the pile, the reduction in the energy transferredto the pile will be minimal, with impedance, Z, being defined as follows :

Z = Ep Ap/c (7.1)

where Ep = Young's modulus of pileAp = cross-sectional area of pile

c = velocity of longitudinal stress wave through the pile

The actual reduction in energy transfer can be measured by dynamic pile testing(Section 8.4) and should be taken into account when taking a final set.

The length of the follower should be limited as far as possible because the longer thefollower, the more difficult it will be to control the workmanship on site. Furthermore,limited site measurements indicated that for follower longer than 4 m, reduction in energytransferred to the pile may occur, even if it is of the same material as the pile section.

Near-shore marine piles in Hong Kong are typically precast prestressed concrete pilesor driven steel tubular piles. Pile driving from a fixed staging is possible for small tomedium-sized piles in waters as deep as 15 m. Alternatively, pile installation may be carriedout with the use of a piling barge or pontoon. Special manipulators and mooring anchoragesare usually required to achieve precise positioning of piles from a barge in deep waters.

7.2.2 Characteristics of Hammers and Vibratory Drivers

(1) General The rating of a piling hammer is based on the gross energy per blow.However, different types of hammers have differing efficiencies in terms of the actual energytransmitted through the pile being driven. The range of typical efficiencies of different typesof hammers is shown in Table 16.

The operational principles and characteristics of the various types of driving equipmentare briefly summarised in the following sections.

(2) Drop hammer. A drop hammer (typically in the range of 4 to 12 tonnes) is liftedon a rope by a winch and allowed to fall by releasing the clutch on the drum. The stroke isgenerally limited to about 1.2 m except for the case of 'hard driving1 into marble bedrockwhere drops up to 2 m have been used in Hong Kong. The maximum permissible dropshould be related to the type of pile material.

The drawback to the use of this type of hammer is the slow blow rate, the difficultyin effectively controlling the drop height, the relatively large influence of the skill of theoperator on energy transfer, and the limit on the weight that can be used from safetyconsiderations. However smaller vibrations associated with the use of drop hammers when

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compared to diesel hammer has led to their wide application in urban sites where vibrationsto adjacent buildings are of concern.

(3) Steam or compressed air hammer. Steam or compressed air hammers areclassified as single-acting or double-acting types depending on whether the hammer fallsunder gravity or is being pushed down by a second injection of propellent. A chisellingaction is produced during driving as a result of the high blow rate. Some single-acting steamhammers are very heavy, with rams weighing 100 tonnes or more.

A double-acting air hammer is generally not suitable for driving precast concrete pilesunless the pile is prestressed.

For maximum efficiency, these hammers should be operated at their designed pressure.The efficiency decreases markedly at lower pressures; excessive pressure may cause thehammer to 'bounce' off the pile (a process known as 'racking') which could damage theequipment.

(4) Diesel hammer. In a diesel hammer, the weight is lifted by fuel combustion. Thehammer can be either single-acting or double-acting. Usually, only a small crane base unitis required to support the hammer. Due to the high noise level and pollutant exhaust gasesassociated with diesel hammers, the use of diesel hammers will be phased out gradually,initially in populated areas.

The driving characteristics of a diesel hammer differ appreciably from those of a dropor steam hammer in that the pressure of the burning gases also acts on the anvil (i.e. drivingcap) for a significant period of time. As a result, the duration of the driving forces isincreased. The length of the stroke varies with the driving resistance, and is largest for harddriving. In soft soils, the resistance to pile penetration may be inadequate to cause sufficientcompression in the ram cylinder of a 'heavy' hammer to produce an explosion, leading tostalling of hammer. In this case, a smaller hammer may be necessary in the early stages ofdriving.

The ram weight of a diesel hammer is generally less than a drop hammer but the blowrate is higher. The actual efficiency is comparatively low (Table 16) because the pressureof the burning gas renders the ram to strike at a lower velocity than if it were to fall freelyunder gravity. The efficiency is dependent upon the maintenance of the hammer.Furthermore, as the hammer needs to exhaust gas and dissipate heat, shrouding to reducenoise can be relatively difficult.

Where a diesel hammer is used to check the final set on re-strike at the beginning ofa working day, results from the first few 'cold' blows may be misleading in that the hammeris not heated up properly and the efficiency may be very low. This source of error may beavoided by warming the hammer up through driving on an adjacent pile.

(5) Hydraulic hammer. In a hydraulic hammer, the hydraulic rams which raise theweight contribute to a higher efficiency and provide flexibility in varying the drop when theground resistance changes significantly. The hammer is less noisy and does not producepolluting exhaust.

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The drawback of this type of hammer is the relatively high capital cost and the needfor a large capacity crane for handling the heavy units. In addition, driving time withhydraulic hammers may be longer than diesel hammers in a hard stratum due to possiblereduction in blow rate at maximum stroke for hydraulic hammers and the increased loss ofhydraulic fluid with the drop height.

(6) Vibratory driver. A vibratory driver consists of a static weight together with apair of contra-rotating eccentric weights such that the vertical force components are additive.The vibratory part is attached rigidly to the pile head and the pulsating force facilitates pilepenetration under the sustained downward force.

The vibratory driver may be operated at low frequencies, typically in the range of 20to 40 Hz, or at high frequencies around 100 Hz (i.e. 'resonance pile driving').

Vibratory hammers are not recommended for precast or prestressed concrete pilesbecause of the high tensile stresses that can be generated.

7.2.3 Selection of Method of Pile Installation

A brief summary of the traditional pile driving practice in Hong Kong is given byMalone (1985).

For displacement piles, two criteria must be considered : bearing capacity anddriveability. Successful pile installation relies on ensuring compatibility between the piletype, pile section, the ground and method of driving.

When choosing the size of a hammer, consideration should be given to whether thepile is to be driven to a given resistance or a given depth.

The force applied to the head of the pile by the driving equipment must be sufficientto overcome inertia of the pile and ground resistance. However, the combination of weightand drop of hammer must be such as to avoid damage to a pile when driving through softoverburden soils. In this case, the use of a heavy hammer coupled with a small drop (longerduration impact and hence larger stress wavelength) and a soft packing is advisable in orderto limit the stresses experienced by the pile head. Conversely, for hard driving conditions,pile penetration will be increased more effectively by increasing the stress amplitude than byincreasing the impact duration.

The weight of the hammer should be sufficient to ensure a final penetration of notmore than 5 mm per blow unless rock has been reached (BSI, 1986). If the hammer is toolight, the inertial losses will be large and the majority of the energy will be wasted in thetemporary compression of the pile. This may lead to over-driving (i.e. excessive number ofblows) causing damage to the pile. Guidance on minimum weights of hammers is given inTable 17.

Other factors which can affect the choice of the type of piling hammer include specialcontract requirements and restrictions on noise and pollution.

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The force that can be transmitted down a pile is limited by a range of factors includingpile and hammer impedance, hammer efficiency, nature of the impulse, characteristics of thecushion and pile-head assembly, and pattern of distribution of soil resistance. If theimpedance is too large relative to that of the hammer, there will be a tendency for the ramto rebound and the input energy to be reflected. Piles with too low an impedance will absorbonly a small proportion of the ram energy, giving rise to inefficient driving. In addition, pileimpedance also has a significant influence on the peak driving stresses. Higher impedancepiles (i.e. heavier or stiffer sections) result in shorter impact durations and generate higherpeak stresses under otherwise similar conditions.

In granular soils, the rate of penetration increases with a higher rate of striking,whereas for stiff clays, a slower and heavier blow generally achieves better penetrationrate.

Commercial computer programs exist for driveability studies based on wave equationanalysis (Section 5.4.3). These can provide information on the stresses induced in the pileand the predicted profile of resistance or blow count with depth.

If a conventional pile driving formula (e.g. Hiley Formula) is used to assess thecriteria for termination of driving, the use of drop hammers or hydraulic hammers (which aremore efficient) could reach the calculated set at greater depths compared to diesel hammersbecause of differences in hammer efficiencies. This highlights the shortcomings of the useof pile driving formulae in determining the required founding level.

A vibratory driver can be effective for the installation of H-piles and open-ended steeltubular piles in granular soils. It is also possible to use this technique to drive the pile to ahard stratum and then use a driving hammer to 'set' the pile. The installation of piles usinga vibrator is not classified as percussive piling under the Noise Control Ordinance (HongKong Government, 1994) and therefore it does not require a percussive piling ConstructionNoise Permit (Hong Kong Government, 1988) during normal working hours. However,caution should be exercised in ensuring that the induced vibrations are acceptable for thesurrounding environment and will not result in undue settlement or damage of adjacentstructures. This may need to be confirmed by field trials where appropriate.

Jetting may be used to install piles into a granular soil but it is generally difficult toassess the disturbance effects on the founding material. This technique is not commonly usedin Hong Kong. Jacking may be considered, particularly for installing piles into clays.

7.2.4 Potential Problems Prior to Pile Installation

(1) General Problems may arise during manufacture and handling of piles, withconcrete piles being more prone to damage than steel piles.

(2) Pile manufacture, filing of concrete during driving may result from sub-standard pile manufacture procedure, particularly where the concrete cover is excessive.Tight control on material quality, batching, casting and curing is necessary to ensure thatsatisfactory piles are manufactured. Lee (1983) noted segregation of concrete in samplesfrom prestressed concrete tubular piles and attributed this to the spinning operation.

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However, the results showed that the design cube strength was not adversely affected.

Recently-cast concrete pile units may crack due to excessive shrinkage as a result ofinadequate curing or due to lifting from the moulds before sufficient strength is achieved.

McNicholl et al (1989a) outlined a quality assurance scheme that has been developedfor prestressed concrete tubular piles in Hong Kong.

(3) Pile handling. Piles may bend considerably during lifting, transportation, stackingand pitching. A bent pile will be difficult to align in the leaders and is likely to be driveneccentrically.

Piles should be lifted by slinging at the prescribed points, and they should not bejerked upwards or allowed to drop abruptly.

Whilst in transit, piles should be adequately supported by blocks to minimisemovements and prevent damage by impact. The blocks between successive layers of pilesshould be placed vertically above the preceding blocks in order to prevent the imposition ofbending forces in the bottom piles.

In stacking piles on site, consideration should be given to the possibility of differentialsettlements between block positions. If the piles are coated with a bitumen layer, particularcare should be taken to avoid damage to the coating by solar heat by means of shading and/orlime washing. The manufacturer's instructions should be strictly adhered to.

A thorough inspection should be made of significant cracks in the piles as delivered.Longitudinal cracking may extend and widen during driving and is generally of greaterconcern than transverse cracking.

If slightly cracked piles are accepted, it is advisable to monitor such sections duringdriving to check if the cracks develop to the point where rejection becomes necessary. Itshould also be noted that when driving under water, crack propagation by hydraulic actionis possible, with water sucked into the cracks and ejected at high pressure.

The criterion for acceptable crack width prior to driving should be considered inrelation to the degree of aggressiveness of the ground and groundwater and the need formaking allowance for possible enlargement of cracks as a result of pile driving. In general,cracks up to 0.3 mm are normally considered acceptable (BSI, 1985), although for bridgedesign, the local practice has been to adopt a limiting crack width of 0.2 mm for buriedstructures.

For concrete within the inter-tidal or splash zone of marine structures, it is suggestedthat the crack width is limited to 0.1 mm (Hong Kong Government, 1992b).

7.2.5 Potential Problems during Pile Installation

(1) General. A variety of potential problems can arise during installation ofdisplacement piles as outlined in the following. Some of the problems that can affect pile

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integrity are summarised in Tables 18 to 21.

(2) Structural damage. Damage to piles during driving is visible only near the pilehead, but the shaft and toe may also be damaged.

Damage to a pile section or casing during driving can take the form of buckling,crumbling, twisting, distortion and longitudinal cracking of steel, and shattering, shearing,cracking and spalling of concrete.

Damage may be caused by overdriving due to an unsuitable combination of hammerweight and drop, and misalignment of the pile and the hammer resulting in eccentric stresses.The hammer blow should be directed along the axis of the pile, but the pile head should befree to twist and move slightly inside the driving helmet to avoid the transmission ofexcessive torsion or bending forces.

Failure due to excessive compressive stress most commonly occurs at the pile head.Tensile stresses are caused by reflection of the compressive waves at a free end and may arisewhen the ground resistance is low or when the head conditions result in hammer rebound, i.e.with hard packing and a light hammer. Damage can also occur when driving from a densestratum into weaker materials. Tensile stresses can result if the pile is driven too fast throughthe transition into the weaker soil. If damage to the head of a steel pile is severe, it may benecessary to have it cut back and an extension welded on.

The driving stresses must not exceed the limiting values that will cause damage to thepile. The following limits on driving stresses are suggested by Standards Association ofAustralia (1995) :

Steel piles : maximum compressive stress > 90% fy

Reinforced concrete piles : maximum compressive stress > 80% fc

maximum tensile stress > 0.8v^"(reinforcement <2%)> vfc" (reinforcement >2%)

Prestressed concrete piles : maximum compressive stress > 80% fc

maximum tensile stress > initial prestress

where fy is the yield stress of steel and fc is the specified grade strength of concrete in MPa.The General Specification for Civil Engineering Works (Hong Kong Government, 1992a)stipulates that the driving stresses in precast reinforced concrete piles and prestressed concretepiles should not exceed one half of the specified grade strength of the concrete, which ismuch more restrictive than the limits proposed by Standards Association of Australia (1995).

Problems at the pile toe may sometimes be detected from the driving records. Thebeginning of easier penetration and large temporary compression (i.e. a 'spongy' response)may indicate the initiation of damage to the lower part of the pile. The blowcount logsshould be reviewed regularly.

Where long slender piles are installed, there is an increased risk of distortion andbending during driving because of their susceptibility to influence of the stress field causedby adjacent piles and excavations.

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Where the bore of prestressed concrete tubular piles is filled with water, Evans (1987)suggested that the hammer impact could generate high pressure in the trapped water andexcessive tensile hoop stresses leading to vertical cracks. In order to detect any dislocationof the pile shoe, the depth of the inner core of each pile should be measured.

A pile with its toe badly-damaged during driving may be incapable of being driven tothe design level, particularly when the piles are driven at close spacings. However, the staticload capacity of such individual piles may be adequate according to load tests due to localcompaction of the upper strata and the creation of a high soil stress at shallow depth due topile driving. The satisfactory performance of any piles during the load test is no guaranteethat the long-term settlement characteristics of the pile group will be acceptable where it isunderlain by relatively compressible soil.

(3) Pile head protection assembly. Badly fitted helmets or the use of unsuitablepacking over a pile can cause eccentric stresses which could damage the pile or the hammer.

The materials used for the dolly and the packing affect the stress waves during driving,depending on whether it is 'hard' or 'soft'. For a given hammer and pile, the induced stresswave with a soft assembly is longer and exhibits a smaller peak stress than if the assemblyis hard. The packing material may be sufficiently resilient initially but could harden afterprolonged use, whereupon it should be replaced. The packing should fit snugly inside thehelmet - too loose a fit will result in rapid destruction of the cushion and hence an undesirableincrease in its stiffness.

The helmet may rock on the pile if the packing thickness is excessive, which couldinduce lateral loads and damage the pile. It is advisable to inspect the pile head protectionassembly regularly for signs of damage.

It should be noted that by manipulation of the packing material, an inadequate pile maybe made to appear acceptable to an unwary inspector in accordance with the pile drivingformula. Only materials with known characteristics should be used for the packing. Pecket al (1974) suggested that wood chips or coiled steel cable are undesirable because theirproperties cannot be controlled.

When a final set is being taken, the packing and dolly should not be new but shouldhave already taken about 500 to 600 blows in order to avoid a misleading set being obtainedas suggested by Healy & Weltman (1980).

(4) Obstructions. Obstructions in the ground may be in the form of man-madefeatures or boulders and corestones.

Obstructions could cause the piles to deflect and break. A steel or cast-iron shoe withpointed or flat ends may be useful, depending on the nature of the obstruction. Where theobstruction is near ground surface, it may be dug out and the excavation backfilled prior tocommencement of driving. If the obstruction is deep, pre-boring may be adopted.Consideration should be given to assessing the means of maintaining stability of the pre-boreand its effect on pile capacity. It should be noted that damaging tensile stresses may resultwhere a precast concrete pile is driven through an open pre-bored hole of slightly smallerdiameter than the pile.

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Experience indicates that 250 mm is the approximate upper limit in rock or bouldersize within the fill or a corestone-bearing profile below which there will be no significantproblems with the installation of driven piles (Hong Kong Government, 1992b).

Alternative options that could be considered include re-positioning of piles, andconstruction of a bridging structure over the obstruction by means of a reinforced concreteraft. Grouting the soil below an obstruction has been suggested by Mackey & Yamashita(1967a) but the effectiveness of the grouting operation may be doubtful.

(5) Pile whipping and verticality. Piles may become out-of-plumb during driving,causing bending and possible cracking. Periodic checks on the verticality of piles should becarried out during driving. The practice of placing wedges between an inclined pile sectionand the next segment to try to correct the alignment should be strongly discouraged.

Where a long slender pile is driven through soft or loose soils, it may be liable to'whip' or wander. This lateral movement during driving may result in a fractionally over-sized hole and affect the skin friction. Pile whipping also reduces the efficiency of thehammer. If the acceptance is based on a final set criterion, it is important to ensure that thereare no extraneous energy losses due to whipping. Failure to do so could result in a pile withinadequate capacity.

Proper directional control and alignment of the hammer and the pile are essential toalleviate the problems. Experience shows that a pointed pile shoe may cause the pile to bedeflected more easily than a flat-ended point.

Broms & Wong (1986) reported a case history involving damage to prestressedconcrete piles due to bending arising from misalignment and non-verticality. A method isproposed to calculate the secondary bending moment that will be induced in a bent pile.

In cases of concern, it may be prudent to cast in or weld on inclinometer ducts formeasurement of pile profile after driving.

Based on results of model tests, Hanna & Boghosian (1989) reported that small kinkscan give higher ultimate load capacity at a larger pile top settlement than that in a straightpile, provided that the pile section is capable of withstanding the bending stresses. For pileswith bends greater than about 10°, it was found that under loading, the increase in stressconcentration and bending may result in overstressing of the adjacent soil and the formationof a hinge, which could lead to a structural failure.

(6) Toeing into rock. A pile is liable to deflect when it encounters the rock surface,particularly where it is steeply-sloping or highly irregular. The pile toe can be equipped witha rock point to give a more uniform stress distribution and minimise the tendency forbending.

A properly reinforced toe is of particular importance when piles are driven into karsticmarble rock surface. Daley (1990) reported his experience with pile driving in marble wherethe toes of H-piles were pointed and the bottom 4 m were stiffened by welded steel plates.Mak (1991) suggested that an abrupt change in stiffness could lead to undesirable stressconcentrations and potential damage, and proposed that a more gradual change in stiffness

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be adopted.

Where a diesel hammer is used, it is advisable to reduce the driving energytemporarily when bedrock is first met to minimize pile deflection. In general, the use of adrop hammer is preferred to help the pile to 'bite' into the sloping rock surface by gentletapping followed by hard driving, as a diesel hammer may be difficult to control at highresistance.

(7) Pile extension. Pile joints could constitute points of weakness if the coupling isnot done properly. The joints should be at least as strong as the pile section. Particular careneeds to be exercised when connecting sections for raking piles.

Steel piles, including H-pile and tubular pile sections, are commonly joined bywelding. It is important that all welding is executed by qualified welders to appropriatestandards (e.g. Hong Kong Government, 1992a). Each weld should be inspected visuallyand, where appropriate, a selection of the welds should be tested for integrity by means ofmechanical or radiographic methods. Alignment of sections must be maintained after weldingand special collars are available as a guide.

In prestressed concrete piles, pile segments are joined by welding together the steelend plates onto which the prestressing bars are fitted by button heads or screws and nuts, andthe reinforcing bars are anchored.

Lengths of precast concrete piles cannot be varied easily. In this case, piles can belengthened by stripping the head and casting on an extension, but this can cause long delaysas the extension must be allowed to gain strength first. Alternatively, special mechanical pilejoints can be used or vertical sections spliced with the use of epoxy mortar dowels. It isimportant to ensure that the abutting ends remain in close contact at all stages of handling anddriving.

Mismatch between the driven section and the extension can occur due to manufacturingtolerances or the head of the driven section having sustained damage in the driving process.It may be necessary to cut off the damaged portion and prepare the end in order to achievea satisfactory weld.

Lack of fit can result in high bending stresses. Joints with a misalignment in excessof 1 in 300 should be rejected (Fleming et al, 1992).

(8) Pre-ignition of diesel hammers. Diesel hammers are prone to overheating andmay pre-ignite when combustion of fuel occurs prior to impact. This leads to a reduction ofthe impact velocity and cushioning of the impact even with a large stroke. Pre-ignition maybe difficult to detect without electronic measurements but possible signs of pre-ignition mayinclude black smoke at large strokes, flames in exhaust ports, blistering paint (due toexcessive heat), and lack of metal-to-metal impact sound. Pre-ignition could considerablyaffect hammer performance and, where suspected, driving should be suspended and thehammer allowed to cool down before re-starting.

In order to function at maximum energy, fuel injected should be adjusted to theoptimum amount and the exhaust set to the correct setting for the appropriate hammer. For

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single-acting and double-acting diesel hammers, the stroke and bounce chamber pressure willgive a reasonably good indication of actual hammer performance. The stroke may bemeasured by attaching a jump stick or barber pole to the hammer for visual inspection or byhigh-speed photographic method.

The hammer performance in terms of energy output per blow (E) may be checkedindirectly by the blow rate. Based on energy considerations, the number of blows per minute(Nb) corresponding to the energy output of a ram weight (W) can be expressed as :

Nb ~ 66vtW/E) (7.2)

where W is in kN and E is in kN-m.

If the measured blow rate is higher than that in the specified energy output, the effectson the energy output should be allowed for in the calculation of the final set. The reductionin energy output may be assumed to correspond to the square of the ratio of Nb to the actualblowcount measured.

It should be cautioned that a hammer in a very poor state of maintenance may havefriction losses of such magnitude that the blow rate will not be an accurate indication ofhammer performance. It is advisable to carry out dynamic load tests to confirm the actualhammer performance, particularly when the use of followers is proposed or when problemsare encountered on site (e.g. premature set at a high level or inability to obtain the requiredset).

(9) Difficulty in achieving set. A method of final set measurement and typical resultsare shown in Figure 45. The supports for the stakes should preferably be at least 1.2 m awayfrom the face of the pile being driven. Difficulties associated with achieving final set havebeen reported in the literature for piles driven into silt, sand and shale (Healy & Weltinan,1980). In these circumstances a hammer with a known impact energy should be used so thatthe actual pile capacity can be assessed. Alternatively pile-head transducers can be installedto measure hammer impact energy.

George et al (1977) suggested that 'wings1 may be fitted to the toes of H-piles in orderto increase the surface area and hence resistance. In principle, where additional steel is tobe welded on near the bottom of a section, it is preferable to have this on the inside of thesection rather than the outside as the latter arrangement may possibly lead to a reduction inskin friction in the long term because of creating an oversized hole.

It should be remembered that the inability to achieve the required set may be attributedto breakage of pile or connections.

For certain geological formations, the pile capacity may increase with time and become'satisfactory. In this case, it may be necessary initially to drive the pile to the minimumrequired penetration and subsequently return to check the final set after a suitable pause.

(10) Set-up phenomenon. There have been a number of documented local casehistories in which piles exhibited an increase in driving resistance when re-driven (Makredes& Likins, 1982; Ng, 1989; Mak, 1990; Lam et al, 1994). In each case, the increase in

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capacity was assessed on the basis of results of repeated dynamic pile tests.

It is possible that the set-up phenomenon is related to dissipation of positive excesspore water pressure generated during driving; alternatively, this may be a result of re-establishment of horizontal stresses on the pile after soil relaxation brought about by pilewhipping. Further work will be required before this effect can be quantified and taken intoaccount in design.

Where a soil exhibits significant set-up, it could lead to problems in achieving therequired penetration length when there are delays to completion of pile installation.Experience has shown that a series of rapidly applied hammer blows using a small drop issometimes successful in 're-starting' a pile after pause.

(11) False set phenomenon. Case histories of problems of false set where thepenetration resistance reduces with time (e.g. Malone, 1977; Thompson & Thompson, 1985)may be associated with the generation of negative pore water pressure during installation orrelaxation of high 'lock-in' stresses in the ground due to the presence of a disturbed zoneassociated with pile driving. The presence of significant cracks in the pile section could alsodampen the stress waves to the extent that false refusal occurs. In some cases, however, theapparent 'relaxation' may not be real in that the difference in penetration resistance is causedby changes in hammer performance. This comment about hammer performance is alsorelevant for apparent set-up as discussed above.

Evans et al (1987) reported that a dynamic load test carried out on a steel tubular piledriven into crushed rock showed a 19% reduction in capacity compared to that estimated uponcompletion of driving. However, tests on other piles in the same site indicated an increasein load capacity.

It is recommended that re-drive tests be carried out on a selection of piles to check forthe possibility of false set.

(12) Piling sequence. Where piles are installed in a large group at close spacing (e.g.saturation piling), consideration should be given to assessing the appropriate piling sequence,paying due regard to the possibility of the ground tightening up and effects of pile uplift.Observations of increase in penetration resistance and increase in SPT.N values with piledriving have been reported by Philcox (1962) and Evans (1987). It is preferable to driveroughly from the centre of a large group and work outwards.

There may be a systematic difference in the pile lengths within a group due to localdensification effects in granular soils. The difference in pile lengths should not be significantas appreciable differential settlements may result. If necessary, extra boreholes may be sunkto confirm the nature of the founding material after pile installation.

For driven cast-in-place piles, there is the possibility of damaging a newly-cast pile asa result of pile driving. Fleming et al (1992) suggested that a minimum centre-to-centrespacing of five pile diameters can be safely employed when driving adjacent to a pile withconcrete less than seven days old. On the other hand, the General Specification for CivilEngineering Works (Hong Kong Government, 1992a) stipulates that piles, including casings;should not be driven within a centre-to-centre distance of 3 m or five times the diameter of

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the pile or casing, whichever is less, from an unfilled excavation or from an uncased concretepile which has been cast for less than 48 hours. In case of doubt, integrity tests may beundertaken to provide a basis for formulating the appropriate guidelines.

(13) Raking piles. Raking piles are comparatively more difficult to install. Whilstraking piles can be driven with a suspended hammer, considerable care is required andsuspended leaders or a piling rig on a crane base may be preferred. Machines whichgenerally carry the pile driving equipment on a long mast will become intrinsically less stablewhen driving raking piles. This is exacerbated by the need to increase the hammer drop inorder to overcome the higher friction involved. Alternatively, the acceptance set may berelaxed where appropriate.

For long piles driven through soft or loose soils, it is possible that a raking pile maytend to bend downward.

Tight control on the alignment of the hammer and the pile is essential. The standardof pile jointing may be affected and the frequency of checking may need to be increased.

(14) Piles with bituminous or epoxy coating. Piles may be coated to minimizenegative skin friction or load transfer to adjacent structures such as underground tunnels. Themanufacturers instructions with regard to the application of coatings, together withrecommendations on the level of protection required, should be adhered to. Extreme careshould be taken to avoid damage to the coating. Pre-drilling may be required to minimizedamage to the coating.

Some guidance on the application of surface protective coating to piles is given in theGeneral Specification for Civil Engineering Works (Hong Kong Government, 1992a).

(15) Problems with marine piling. Problems that may arise with marine piles includedifficulties with piling through obstructions such as rubble mounds, necking, buckling andinstability associated with piling through water or through a thick layer of very soft marinedeposit and the need for pile extension over water.

A relatively stable working platform is essential for pile installation. Piles may bedriven from a temporary staging, spudded pontoon or floating craft. The latter will besubject to tidal effects and regular adjustments may be necessary to maintain a pile in line.It is generally inadvisable to use a drop hammer on a . floating. craft because of potentialproblems of directional control.

There is the likelihood of damage to precast concrete piles driven from a barge,especially at exposed sites. Under certain circumstances, pile driving from a barge may beacceptable for relatively protected sites, particularly where steel piles are to be used. Largepiling barges should be used to minimise the possibility of piles being damaged due to bargemovements.

Gates or clamps may be necessary to assist alignment and facilitate pile extension.Care needs to be exercised in the design of such devices to maintain pile position andtolerances, particularly in the case of raking piles as there is a tendency for the pile to shiftlaterally. This, coupled with the weight of the hammer and the free standing portion of the

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pile, may lead to damage of the gates.

For marine piles, it is important to ensure that adequate bracing to pile heads, in twodirections at right angles, is provided immediately after installation to prevent the possibilityof oscillation in the cantilever mode due to current and wave forces.

Typical case histories of marine piling in Hong Kong are reported by Construction andContract News (1983) and Hazen & Horner (1984).

Practical aspects and considerations related to maintenance of marine piles in serviceare discussed in Hong Kong Government (1992b).

(16) Driven cast-in-place piles. For top-driven tubes with a flat or conical cast ironshoe, the shoe is liable to be damaged by an obstruction and it should be checked duringdriving by sounding with a weight.

For a casing driven by an internal drop hammer, it is important that the dry concreteplug at the base is of the correct consistency. Otherwise, driving may not cause the plug tolock in the casing, leading to ingress of soil and water. As a general guideline, thewater/cement ratio should not exceed 0.25 and the plug should have a compacted height ofnot less that 21A times the pile shaft diameter. Heavy driving may result in bulging of thecasing or splitting of the steel if the plug is of inadequate thickness. Fresh material shouldbe added after prolonged driving (e.g. two hours of normal driving and one hour of harddriving) to ensure that the height of the plug is maintained.

The relatively thin bottom-driven steel casing is liable to collapse when piles are driventoo close to each other simultaneously, and can result in loss of the hammer. The risk of thishappening is increased when piles are installed within a cofferdam where there may be highlocked-in stresses in the ground.

Problems could arise during the course of concreting of driven cast-in-place piles(Section 7.3.5(2)).

A useful discussion on the construction control of driven cast-in-place piles is givenby Curtis (1970).

(17) Cavernous marble. In cavernous marble, karst features that could give rise todesign and construction difficulties include pinnacles, solution channels and slots, cliffs,overhangs, cavities, rock slabs or blocks, collapsed or infilled cavities. Potential problemsassociated with driven piles include large variation in pile lengths, pile deflection, local over-stressing due to inclined rock surface, inability to penetrate thin slabs which may be underlainby weaker materials, damage to pile toe, uncertain effects of driving and loading of a pilegroup on cavity roofs, bending and buckling of piles in the overburden and the possibility ofsinkhole formation as a result of collapse of cavities induced by pile driving (Houghton &Wong, 1990).

Due to the uncertainties in ground conditions in the form of irregularities of the rocksurface and the presence of cavities, it is common in Hong Kong to continue with 'harddriving' after the pile has keyed into rock; The aim is to facilitate penetration through thin

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roof slabs that may be present. However, overdriving leading to toe damage and bendingshould be avoided and a heavy section is essential to prevent buckling during driving. Bettercontrol may be exercised by using a drop hammer for hard driving in conjunction with astrengthened pile shoe, which should preferably be pointed.

Re-driving tests should be carried out because of the possibility of damage to thefounding stratum caused by hard driving which may affect adjacent piles previously installed.

A case history of piling in faulted marble is described by Yiu & Tang (1990).

7.2.6 Potentially Damaging Effects of Construction and Mitigating Measures

(1) Ground movement. Ground movements induced by the installation ofdisplacement piles causing damage to piles already installed have been reported in Hong Kong(Short & Mills, 1983). Significant ground heave is possible and could lead to pile uplift.A useful summary of the mechanism of ground movements is given by Hagerty & Peck(1971). Premchitt et al (1988) reported ground heave of 150 mm near each prestressedconcrete tubular pile after driving through marine clay and clayey alluvium. Siu & Kwan(1982) observed up to 600 mm ground heave during the installation of over 200 driven cast-in-place piles into stiff silts and clays of the Lok Ma Chau Formation. Mackey & Yamashita(1967b) stated that problems of foundation heave due to construction of driven cast-in-placepiles had been encountered where the ground consisted of colluvial decomposed granites, butthat this was rare with insitu decomposed rock.

Uplift of piles can cause unseating of an end-bearing pile, leading to reduced stiffnessand capacity, or breaking of joints and/or pile shaft, particularly if the pile is unreinforcedor only lightly reinforced.

The problem of ground heave and pile uplift may be alleviated by pre-boring.Alternatively, a precast pile may be redriven after it has been uplifted. Experience has shownthat it may not be possible to redrive uplifted piles to their previous level and that a similarset may be acceptable at a slightly higher level. As driven cast-in-place piles cannot be easilyredriven once concreted, Cole (1972) suggested the use of the 'multi-tube'technique wherebythe temporary liners for all the piles within eight diameters of each other are installed firstand reseated prior to commencement of concreting. The technique was found to be effectivein reducing pile uplift. However, it requires careful planning and the availability of a numberof temporary liners. These two elements may render the technique costly and less attractiveto large piling projects.

Uplift trials may be carried out during load test to assess the effect of uplift on pileperformance (Hammon et al, 1980).

Ground movements induced by driving could affect retaining structures due to anincrease in earth pressures. Lateral ground movements can also take place near river banks,on sloping sites, at the base of an excavation with an insufficient safety margin against basefailure or near an earth-retaining system (e.g. sheetpiles) with shallow embedment. Theeffect of such potentially damaging ground movement on a pile depends on the mode ofdeflection, i.e whether it behaves as a cantilever with high bending stresses or whether it

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rotates or translates bodily. In addition, twisting of a pile may induce undesirable torsionalstresses.

Levelling and surveying of pile heads and possibly the ground surface should beinstigated if significant ground movement is expected or suspected. Consideration should begiven to assessing the optimum piling sequence and the need for pre-boring. The spacing ofthe piles could also be increased to minimise the problem. The sequence of driving does notappear to have an appreciable effect on the total amount of uplift but it may be varied so thatany uplift is distributed in a manner more favourable to the structure. Alternatively, a small-displacement pile solution may be adopted. In extreme cases, the risk of damage to sensitivestructures could be minimised by constructing a 'relieving' trench filled with compressiblematerial, although the effectiveness of such proposals will need to be confirmed by fieldtrials.

It should be borne in mind that pile top deflection cannot be regarded as the sole factorin assessing the integrity of a displaced pile. Tools that can be used for investigation includeintegrity tests, re-driving, dynamic and static load test, and exhumation of piles for inspectionwhere practicable. As a rough guide, Bjerrum (1957) recommended that H-piles having aradius of curvature of less than 366 m after driving will be unacceptable for carrying load.On the other hand, Broms (1963, 1984) described methods to determine the reduced capacityof bent piles.

It is generally inadvisable to attempt to correct laterally displaced piles by jacking atthe pile heads as this could lead to failure of the section in bending.

(2) Excess pore water pressure. Siu & Kwan (1982) and Lam et al (1994) reportedobservations of generation of positive excess pore water pressure during pile driving. Thedissipation of the excess pore pressures could lead to the phenomenon of pile set-up (Section7.2.5(10)).

In soft clays and marine mud, the dissipation of excess pore pressures may give riseto negative skin friction (Lumb, 1979). Small-displacement piles with vertical drains attachedmay be considered to minimise this effect in extremely sensitive clays.

Where piles are driven on a slope, the excess pore pressure could result in slopeinstability. Where soft clays are involved, the induced pore pressures may lead to hydraulicfracture of the ground giving rise to crack formation. This may in turn increase the capacityfor infiltration.

In soft sensitive clays, the effects of excess pore pressure and remoulding may resultin a significant reduction in shear strength. This will be important in the case of piles forabutments where the clay will induce horizontal loading and hence stresses in the pile.

(3) Noise. Percussive piling is inherently noisy and the operation is subject to theNoise Control Ordinance (Hong Kong Government, 1994). The Ordinance stipulates thatpercussive piling requires a Construction Noise Permit (Hong Kong Government, 1988),which may restrict the permitted period of operation to three hours per day. Usefulbackground discussions on the nature of various types of noise, the methods of measurementand means of noise reduction are given by Weltman (1980a) and Kwan (1985). Sources of

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noise from percussive piling operations include radiation of noise from the hammer exhaustand impact of hammer. Shrouds are normally used for noise control which can result inreduced hammer efficiency and increased cost. Cockerell & Kan (1981) suggested that noiseradiated from the pile itself may be comparable to that from the hammer and exhaust suchthat even an effective shroud fitted over the hammer will reduce the total noise by only about50%.

It should be noted that bottom-driven piles will generate less noise than piles whichare driven at the top.

Table 22 summarises the typical range of noise levels associated with different typesof piles and the use of related construction equipment based on local measurements in about200 construction sites in Hong Kong.

(4) Vibration. The prediction of the vibration level which may be induced for aparticular combination of plant, pile and soil condition is fraught with difficulties The natureand effects of ground-borne vibrations caused by piling are discussed by Head & Jardine(1992).

Vibration due to pile driving (or installation of a temporary casing for replacementpiles) may lead to compaction of loose granular soils or loose voided fill and cause theground surface or utilities to settle (O'Neill, 1971; Esrig et al, 1991). In addition, dynamicstresses will be induced on underground utilities and structural members of buildings. Theresponse of different forms of construction will vary and certain structural details may leadto a magnification of the vibration effect (Heckman & Hagerty, 1978).

The most commonly used index for assessing the severity of vibration is the peakparticle velocity, ppv. As the problem of wave propagation and attenuation is complex, themost practical approach is to make reference to results of field monitoring of similarconstruction in similar ground conditions. Figure 46 summarizes some of the publisheddesign lines derived from monitoring results. Luk et al (1990) reported results of vibrationmonitoring carried out during driving of prestressed concrete tubular piles in the Tin ShuiWai area. They concluded that the following equation proposed by Attewell & Farmer (1973)can be used as a conservative upper bound estimate of the free-field vector sum peak particlevelocity, ppv (in mm/sec) :

(7'3)

where k = constantE = driving energy in joules

Ah = horizontal distance from the pile axis in metres

The above recommendation may be used with a k value of 1 .5 as a first approximationbut it will be more satisfactory to develop site-specific correlations. Limited monitoringresults in Hong Kong suggest that the upper limit can be refined to correspond to a k valueof unity for precast concrete piles, and a k value of 0.85 for H-piles.

The transmission of vibration energy from the pile to the soil is controlled by pile

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impedance, and during wave propagation in the ground the vibration attenuation is influencedby the damping characteristics of the soil, wave propagation velocity and vibration frequency(Massarch, 1993). These factors are not directly considered in most empirical relationships.

In Hong Kong, there is no official legislation or code of practice on vibration control.However, for the protection of Mass Transit Railway structures, the BOO Practice Note forAuthorized Persons and Registered Structural Engineers (BOO, 1977) stipulates that the ppvat any railway structures resulting from driving or extraction of piles or other operationswhich can produce 'prolonged' vibration shall be limited to 15 mm/sec. The allowable ppvand pseudo-dynamic ground movements have been considered in a number of overseas codesalthough most of the recommendations have not been drawn up specifically for groundvibrations induced by piling. The behaviour is strongly affected by local conditions andextreme caution needs to be exercised in extrapolating these criteria.

As a general guideline, experience in Hong Kong has shown that a limiting ppv of 10to 12 mm/sec is generally acceptable for major underground utilities and water retainingstructures, whereas the corresponding values for buildings are in the range of 15 to25 mm/sec which are likely to be conservative. For water retaining structures and certainutilities including water mains, an additional criterion in terms of a limiting dynamicdisplacement (e.g. 200 />cm in general and 100 /xm for water retaining structures) may beimposed as appropriate. For buildings of historical significance, the limiting ppv valuesrecommended in various overseas codes are in the range of 2 to 3 mm/sec. Limitedexperience in Hong Kong indicates that a ppv of 6 to 8 mm/sec can be acceptable. Inprinciple, consideration should also be given to the duration over which the peak vibrationtakes place in assessing the limiting ppv values.

Due to the complexities involved, it may not always be appropriate to rely on theabove generalised guidelines. It is advisable that each site is assessed on its merits, takinginto consideration the existing condition of the structures, possible amplification effects andpotential consequence of failure. In critical cases, it would be advisable to carry out testdriving combined with vibration monitoring to assess the potential effects and define a moreappropriate and realistic limit on acceptable piling-induced vibration. In determining theacceptable threshold limits, consideration may also be given to the dominant frequency ofexcitation and the duration of vibration (Selby, 1991). It has been found that larger ppvvalues will be acceptable at a higher frequency of vibration (Head & Jardine, 1992). Also,the limiting ppv value may be lower for continuous vibration than for intermittent vibration.

Where significant vibration is envisaged or where the surrounding structures aresensitive (e.g. pressurised water mains or computers in buildings), it will be prudent to carryout vibration monitoring during test driving and installation of trial piles. A settlement surveyis also helpful in monitoring settlement resulting from pile driving. Based on the initialmeasurements, the suitable course of action, including the need for continual monitoringduring site works, can be assessed. A comprehensive dilapidation survey of the adjacentstructures with good quality photographs of sensitive areas or existing defects should becarried out prior to commencement of the works. A case history on an engineered approachin assessing and designing for potential vibration problems is described by Grose & Kaye(1986).

Measures which may be considered to reduce piling vibration include :

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(a) control of number of piles being driven at any one time,

(b) pre-boring,

(c) change of piling system,

(d) 'active' isolation - screening by means of a wave barrier(e.g. trench, air cushion) near the energy source, and

(e) 'passive' isolation - screening by means of a wave barriernear the affected structures.

The effectiveness of a wave barrier is related to the amplitude and energy of thewaves, and the barrier dimensions. A design method is put forward by Wood (1967). Liao& Sangery (1978) discussed the possible use of piles as isolation barriers. The effectivenessof the barriers should be confirmed by field trials as theoretically it is possible foramplification to take place for a certain combination of conditions.

Provided the accepted method of installation is proved by instrumented test driving,the sequence of piling may be stipulated to have the piles driven in a direction away from thesensitive structures so that stresses are not built up.

7.3 INSTALLATION OF MACHINE-DUG PILES

7.3.1 Equipment

(1) General. Machine-dug piles may be founded in saprolite, or bearing on orsocketed into fresh to relatively unweathered rock (i.e. Grades I to III).

(2) Large-diameter bored piles. The range of drilling equipment developed forconstructing large-diameter bored piles has been reviewed by Stotzer et al (1991). Two maintechniques can be recognised on the basis of the method of excavation and means of groundsupport. The 'casing-support' technique involves excavation by a high table rotary rig orgrabs and chisels within a steel casing which is advanced progressively with the use of anoscillator or vibrator. With the advent of hydraulic rigs with the ability to insert tools overprotruding casing, rotary methods are faster than grabs and chisels in most soil conditions.Alternatively, a proprietary system involving the use of a pneumatically-powered 'swinghead'may be adopted, which can be time-consuming but would be particularly useful for pilingon a steeply-sloping site. Where excavation is carried out beyond the casing, the bore willneed to be supported by an excess head of water (Au & Lo, 1993) or, where necessary, bydrilling fluids such as bentonite slurry.

The ' slurry-support1 technique involves excavation of a shaft under a drilling fluid withthe use of a reverse-circulation drill, rotary auger or rotary drilling bucket. In less weatheredzones, a reverse-circulation drill incorporating rock roller bits may be used. Alternatively,a core barrel can be employed using air or water circulation. Recently, a multi-head hammerdrill incorporating down-the-hole hammers has been introduced to Hong Kong which, withproper control measures implemented, can result in increased drilling rates. Barrettes may

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be formed in short trenches using conventional diaphragm walling equipment of grab andchisel. A milling machine powered by down-the-hole motors with reverse mud circulationcan also be used to form barrettes in less weathered rock.

Sellouts may be formed with the use of a reverse circulation drill incorporating anunder-reaming head or by 'offset-chiselling1 using an inclined chisel guide.

(3) Mini-piles. These piles are usually constructed with the use of rotary direct-circulation drilling, although reverse-circulation drilling equipment are also available. The'duplex system9 is sometimes employed where the rod and the casing are advanced togetherand the return flush travels up within the casings, thereby minimising soil erosion along theshaft. Down-the-hole hammers may be used to break up boulders. Alternatively, a down-the-hole hammer incorporating a reaming tool may be used, particularly in poor groundconditions.

(4) Continuous-flight auger (cfa) piles. These piles are installed by drilling with arotary continuous-flight auger to the required depth which is generally less than 30 m. Afterreaching the required depth, grout (or highly workable concrete in larger diameter piles) ispumped down the hollow stem and fills the void as the auger is slowly withdrawn, with orwithout being rotated. The walls of the borehole are continuously supported by the spiralflights and the cuttings within them. On completion of grouting, reinforcement cage up to20 m long is pushed into the grouted hole.

7.3.2 Use of Drilling Fluid for Support of Excavation

(1) General. Construction of bored piles involves shaft excavation and adequatesupport must be provided to prevent bore collapse and minimise the effects of stress reliefand disturbance of the surrounding ground. Some loosening of the soils is inevitable duringexcavation but if the degree of disturbance is uncontrolled, the effect on pile performancemay be significant and variable.

Drilling fluids may be used to provide bore support in an unlined hole. This may bein the form of bentonite slurry, polymer mud or water where appropriate. The use of drillingfluid to support pile excavations in a steeply-sloping site should be viewed with caution anda sufficient length of lead casing should be advanced where possible to minimise the risk ofhole collapse due to differential earth pressures.

Because of the larger volume of drilling fluid needed to be treated prior toreintroduction into the bore, all reverse circulation drills require control of the suspensionsystem.

(2) Stabilising action of bentonite slurry. The successful use of bentonite slurry asa means of excavation support relies on the tight control of its properties. A comprehensivesummary of the stabilising action of bentonite slurry and polymer fluids is given by Majano& O'Neill (1993).

The inherent characteristics of bentonite slurry are its ability to swell when wetted, itscapability in keeping small sediments in suspension, and thixotropy, i.e. it gels when

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undisturbed but flows when it is agitated.

The slurry penetrates the walls of the bore and gels to form a filter cake which actsas a sufficiently impervious diaphragm to allow the transmission of hydrostatic slurrypressure. To ensure bore stability, the hydrostatic pressure of the bentonite slurry must begreater than the sum of the water pressure and the net pressure of the soil.

(3) Testing of bentonite slurry. The essential properties of bentonite slurry includedensity, viscosity, fluid loss, sand content, pH and filter cake thickness. Conventionalrequirements on the shear strength of the slurry developed for oil drilling purposes are of lessrelevance to civil engineering works. Generally speaking, density, viscosity and fluid lossare the more relevant control parameters for general piling works whereas pH is a usefulindicator on the degree of contamination of the slurry, although experience exists of poor pileperformance where the sand content or the filter cake thickness is excessive. It is advisableto adopt a flexible approach in determining the range and extent of compliance testingrequired for each site, which should be reviewed as the works proceed. Although thepressure on site for concreting is inevitably great, it is important to ensure compliance of thebentonite slurry properties with the specification requirements, as otherwise the integrity orthe resistance of the pile or both may be compromised.

Bentonite slurry will become contaminated with soil sediments during excavation.Limits on slurry properties are normally stipulated for slurry as supplied to the pile, and forbentonite immediately prior to concreting. A useful background discussion can be found inHutchinson et al (1974).

There is no universally agreed specification on the limits on the properties of bentoniteslurry. It is necessary to take into account the origin and nature of the bentonite whenassessing the limits on the properties. Typical specifications published in the literature areshown in Table 23. Some local contractors have adopted more stringent control on propertiesof bentonite. An example of such specification on properties of bentonite is shown inTable 24.

(4) Polymer fluid. Polymer fluids have been used to maintain bore stability duringexcavation as an alternative to bentonite slurry (Corbet et al, 1991). The advantages ofpolymer fluids include simpler site logistics, rapid hydration, less requirement for storage,less disposal problems, inertness to cement and absence of a filter cake. However, polymerscan be difficult to mix; the shearing action must be sufficiently high to disperse the polymersbut not so great as to break down the polymers. In addition, polymer fluid can be susceptibleto becoming wet and forming a slime.

Beresford et al (1987) discussed the testing of polymer fluid and suggested acceptancecriteria for the results. Experience with the use of polymer fluid has been very limited inHong Kong.

7.3.3 Assessment of Founding Level and Condition of Pile Base

For piles bearing on rock or socketed in rock, pre-drilling is necessary to establish therequired founding level. Cores (minimum of NX size) are normally taken to at least 5 m or

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three times the average pile diameter below the proposed pile base level, whichever is larger,except for sites underlain by marble, in order to prove the nature of the founding material.The acceptable values of index parameters, such as total core recovery, unconfinedcompressive strength (or point load strength), RQD and the nature of discontinuities belowthe founding level etc., must be determined in relation to the design method. The chippingsshould be inspected carefully to confirm the nature of the material when the proposedfounding level is reached. Comments have been given in Section 5.5.3(4) on the potentialshortcoming in the use of total core recovery or RQD as the sole means of determiningsuitable founding level. More than one criterion should be used, where possible, to evaluatethe required founding level.

In principle, geophysical testing techniques can be used to assess the appropriatefounding level. In practice, such indirect techniques may not be sufficiently reliable fordetailed foundation design.

For large-diameter bored piles bearing on rock, it is common for core sampling to bestipulated for a selection of contract piles. This involves the retrieval of minimum 100 mmdiameter cores through the concrete shaft which may be extended to, say, a minimum of600 mm or a distance of half a pile diameter below the base in order to assess the conditionof the pile/rock interface and confirm the nature and state of the founding material. If coresare taken only to assess the base interface, NX size core taken through a void former cast intothe pile would generally be adequate. The frequency of retrieving cores may vary betweensites, depending on the contractor's experience and the designer's confidence. As generalguidance, it is suggested that a minimum of one to two cores should be taken for every 100piles, but judgement should be exercised for individual projects, taking into account thecomplexity of ground conditions, the problems encountered during pile construction, the scaleof the work, etc.

Where defects at the pile base, such as debris and mud, are revealed, the interfacewould need to be flushed out and the void grouted. Several holes may be required tofacilitate the flushing of the debris. Further cores should be taken to verify the effectivenessof remedial grouting in each pile.

For rock-socketed piles, base coring may be supplemented with shaft interface coringwhere there are doubts as to the effectiveness of bonding or adequacy of the concrete.Successful sampling of interface cores has been reported in the literature (Holden, 1984)where the strength of the rock is comparable to that of concrete. However, the operation isvery costly, delicate and requires considerable skill to successfully recover representativecores. Alternatively, the adequacy of the bonding can be investigated by means of a load teston an instrumented pile.

For piles founded in saprolite, Standard Penetration Tests are normally carried out toenable the required founding level to be assessed. Plate bearing tests (Sweeney & Ho, 1982)or pressuremeter tests (Chiang & Ho, 1980) can also be used to characterise the ground anddetermine design parameters.

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7.3.4 Potential Problems during Pile Excavation

(1) General A range of potential problems can arise during the installation of boredpiles and these are outlined in the following. Some of the problems that can affect thestructural integrity of piles are summarised in Table 25.

(2) Bore instability and overbreak. Overbreak arises where there are local collapsesof the walls of the bore resulting in cavities. These cavities, particularly if they are water-filled or slurry-filled and concealed behind a temporary casing, pose a potential risk ofcontamination of the concrete when the casing is extracted. Surging of the casing should beavoided as this will increase the likelihood of ground loss and hence settlement. The profileof the excavation and the degree of overbreak may be assessed approximately with the useof a mechanical or sonic calliper measuring device. However, it is not possible to calliperoverbreak which is concealed by a temporary casing. Alternatively, the profile of excavationcan be roughly estimated by back-calculating from the volume of concrete used inconstructing the pile.

It is important to ensure that there is a sufficient excess hydraulic head within thecasing against base blowing and to prevent shaft instability where excavation proceeds belowthe casing. In the case where water is used to support an excavation below the casing,consideration should be given to the risk of bore instability when the excess water headreduces due to breakdown of pumps or seepage into the ground between shifts, e.g. overweekends.

Rapid withdrawal of a drilling bucket or hammer grab during pile excavation shouldbe avoided as this may give rise to undercutting beneath the casing as well as a 'piston effect'resulting in significant reduction in pressure and bore collapse. Specially-designed bucketswhich have a by-pass arrangement to allow the flow of bentonite fluid to take place to reduceany severe damage to the wall of the pile shaft (Fleming and Sliwinski, 1977) may be used.

(3) Stress relief and disturbance. Pile bore excavation will result in stress relief ofthe ground. Stroud & Sweeney (1977) observed from a trial diaphragm wall panel that at anapparent excess slurry head of 1.5 m, completely weathered granite exhibits considerableswelling and ground loss and settlement. A minimum excess slurry head of 3.5 m wasspecified for the diaphragm wall for the Hongkong & Shanghai Bank Building (Nicholson,1987). Excessive swelling and loosening could also affect the stiffness and capacity of piles.

Where a full length temporary casing is used, the process of oscillating or vibratingthe casing may cause disturbance to the soil structure. Excavation below the casing or thetendency for seepage flow to occur towards the bottom of the excavation will lead to furtherdisturbance and loosening of the soil in the pile shaft by stress relief or seepage forces.

Where the piles are bearing on rock, the above disturbance effects may not be ofsignificance. However, for piles founded in saprolite, the effects should be considered in theassessment of the available shaft capacity. The stress relief and disturbance effects can beminimised by maintaining a sufficient excess hydraulic head at all times or ensuring that thecasing is always advanced to beyond the excavation level.

(4) Obstructions. With reverse-circulation drills or down-the-hole tools, the presence

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of obstructions can generally be overcome relatively easily. It should be noted however thatthe use of the airlift technique as a means of flushing (which relies on the suction effect dueto the difference in density between the air-water mixture and the surrounding fluid) requiresa hydraulic head of about 10 m and therefore shallow obstructions cannot be dealt withreasonable performance by reverse-circulation drills. This problem can be alleviated by usingsuction pump together with a down-the~hole hammer drill. With the casing-support method,chisels are usually used. For obstructions and boulders with a sloping surface, it should beborne in mind that the chisel may skid sideways upon impact and could damage the steelcasing.

For major obstructions, a possible option will be to backfill the bore with spoil orpreferably lean mix and excavate the obstruction with the use of an oversized casing. Smalldiameter drillholes may also be sunk to perforate the obstruction to facilitate subsequentbreaking up by a chisel. However, careful consideration needs to be given to the possibilityof contamination of the bentonite slurry by the cement in the lean mix.

Manual excavation has sometimes been resorted to for relatively shallow excavationsabove the water table. For obstructions at depth, the extent of ground treatment required tominimize the safety hazard and effects of dewatering needs to be carefully assessed prior toconsideration of manual excavation.

(5) Control of bentonite slurry. The quality and level of the bentonite slurry must bekept under tight control during bore excavation. The bentonite should be mixed with freshwater by means of a properly-designed mixer and left for a sufficient time to achieve effectivehydration. In the presence of seawater or in areas affected by saline intrusion, suitableadditives may be necessary to maintain the properties of bentonite slurry as a stabilising fluid.

Contamination by clay minerals (e.g. in marine mud), particularly in the form ofcalcium or aluminium ions, could promote ion exchange with the slurry such that the filterproperties are markedly changed. In this case, the filter cake could become thicker and havea far higher fluid loss, which can cause the gel structure of the slurry to collapse leading tobase instability. Contamination by cement will result in similar effects together with a largeincrease in the pH value. The increase in filter cake thickness may not endanger borestability but could affect the mobilised skin friction as the filter cake may not be effectivelyscoured and removed by the concrete. The presence of a filter cake will create a lubricatingsurface and prevent the cement milk from penetrating the disturbed soil.

The pH of the slurry should be kept in the alkaline range but this may be influencedby the minerals present in the water and the soil. In particular, organic soils could cause thebentonite to become thin and watery, and cease to perform its functions (Reese & Tucker,1985).

Bentonite slurry is liable to 'run away' in very permeable (e.g. ks > 10"2 m/s) strata.The nature of some reclamation fill may pose a risk of sudden loss of bentonite leading tobore collapses. At the Cityplaza development at Taikoo Shing in Hong Kong, it was foundnecessary to construct a pre-trench filled with lean-mix concrete to alleviate this potentialproblem (Craft, 1983). Similar problems of risk of sudden loss of bentonite can arise incavernous marble, landfill sites and in the vicinity of underground utility service pipes orducts.

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Nicholson (1987) reported results of piezometers which show outward flow of waterfrom a diaphragm wall trench at the end of a day's excavation and restoration of theequilibrium groundwater level by the following morning. It was conjectured that where theexcess bentonite head is insufficient to prevent excessive swelling of some of the weatheredgranites, the inward movement coupled with the continual raising and lowering of the grabcould cause disturbance or shaving-off of the filter cake, which re-developed overnight. Itis therefore important to maintain a sufficient excess bentonite head and use a bentonite slurrywhich forms a filter cake rapidly. It may be possible that the use of reverse circulationdrilling may lead to less disturbance of the filter cake compared to that of a grab, leavingpotentially a relatively smooth bore profile along the shaft.

The built-up of filter cake thickness varies with the square root of time (Nash, 1974).Hence a pile bore should not be left open for an excessive period of time as this could leadto a thick filter cake developing on the sides of the excavation. Careful consideration shouldbe given to the programming of excavation and concreting.

(6) Base cleanliness and disturbance of founding materials. Debris accumulated atthe base of a pile is undesirable as this may lead to intermixing and inclusions in the concrete.Debris may comprise soft and loose sediments that settle to the base after completion ofexcavation. Alternatively, foreign materials could be deposited accidentally into the pile.It will be prudent to ensure that a sufficient projection of the temporary casing (say, 1 m) isleft above ground level and that empty bores are properly covered.

The final cleaning of the pile base may be done with the use of a cleaning bucketfollowed by airlifting (Sliwinski & Philpot, 1980). The use of a skirted airlift in which debriswould be drawn in over a larger area may be more effective (Fleming et al, 1985). On someoccasions, the reverse-circulation drill has been used for this purpose. Opinions differ as tothe effectiveness and potential disturbance between the use of an airlift pipe and the reverse-circulation flush, particularly in weathered rocks which may be susceptible to disturbance ordamage of the bonding inherent in the grain structure. Thorough base cleanliness may bedifficult to achieve in practice, particularly with raking piles. If base cleaning is not doneproperly, potential problems including plastering of the filter cake and presence of largepieces of debris at the pile base may occur.

Even if the base is free from significant debris, the soil below the base may bedisturbed and loosened as a result of digging, stress relief or airlifting (Section 7.4.3(4)).Special techniques may be adopted to consolidate and compact the loosened soil. Theseinclude pressure grouting with the use of a stone fill pack (Tomlinson, 1994) or tube-a-manchette (Sherwood & Mitchell, 1989). In addition, shaft grouting may be carried out toenhance the shaft stiffness and capacity (Morrison et al, 1987). However, Mojabi & Duffin(1991) reported that no significant gain in skin friction was achieved by shaft grouting insandstone and mudstone. Experience with such construction expedients is limited in HongKong.

Rock-socketed piles are liable to base-cleanliness problems arising from fine rockmaterials. If the debris is not removed properly, a 'soft toe1 may form at the base of the pile.Fresh concrete may also force the base debris up the socket wall thereby reducing the sideshear resistance in the lower region of the socket.

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(7) Position and verticality of pile bores. The position of pile bores should bechecked as piles significantly out of position may necessitate a reassessment of the pile capcarrying capacity. Non-verticality of a pile bore will induce additional bending and maynecessitate extra reinforcement if it is seriously in error. It is common practice in HongKong to routinely check the verticality of the casing to ensure acceptable verticality of thepile bore. This could involve the use of a dummy reinforcement cage, or a sonic ormechanical calliper device.

For barrettes, it is important to ensure that a guide wall of sufficient depth isconstructed to guide the grab.

For piles installed close to tunnels or which are required to be constructed to very tighttolerances (e.g. piles for top-down deep excavation), precautions may need to be adopted inthe construction including the use of precise instruments for control and verification of theverticality (Triantafyllidis, 1992).

(8) Vibration. Vibration may be caused when a temporary casing is vibrated into theground. The problems of excessive vibration are discussed in Section 7.2.6(4). Where avibrator is used, adjusting the operating frequency of the vibrator may in some cases help toreduce the level of excited ground vibrations.

(9) Sloping rock surface. The installation of temporary casings to obtain a seal inrock may be fraught with difficulties where the rock surface is sloping. A possibleconstruction expedient was described by Mckenna & Palmer (1989) involving the use of weakmass concrete to plug the gap between the casing and the rock surface followed by furtherdrilling into rock after the concrete has hardened.

(10) Inspection of piles. For rock-socketed piles resisting lateral loads and where thecasing can obtain a seal to prevent ingress of groundwater, the socket should be inspected toexamine if there are any adversely oriented joints which may affect the design. It will alsobe good practice to check the condition and dimension of a belled pile by descent in dryconditions. Safety precautions to be taken during the descent must be carefully considered.Reference may be made to BSI (1978) and the Guidance Notes on Hand-Dug Caissons (HongKong Institution of Engineers, 1981).

The use of a closed-circuit television to inspect a rock socket in lieu of inspection bydescent may be considered provided the designer is satisfied that this technique is sufficientlyreliable. The use of television or video camera for inspecting piles in clays can be unreliableand is not recommended because the clay may be smeared by the drilling tool.

Machine bored piles constructed under water have also been inspected by divers(Mckenna & Palmer, 1989).

With the advent of mechanical under-reaming tools which can be calibrated at theground surface, checking of bell-out dimensions using the above methods may gradually berelaxed.

(11) Recently-reclaimed land. In the case of piles constructed through a recentreclamation where marine mud may be trapped and disturbed with excess (possibly artesian)

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pore water pressure, a stable bore may be difficult to achieve. Raised guide walls ordewatering or both where practicable, or the use of a full length casing through the soft areasas appropriate, may be required to prevent bore collapse.

(12) Bell-outs. It is generally difficult to verify the dimensions and ensure thoroughcleaning of a bell-out. In piles constructed under water the use of divers may be consideredbut the reliability of such a procedure as quality control may be questionable. In water depthsin excess of about 10 to 15 m, this will become a very slow process as diver decompressionwill be required.

As a minimum, the vertical displacement of the drill string should be measured if amechanical under-reamer is used to form a bell-out. Similarly, a reference mark on the chiselwire may be used in the case of offset-chiselling to check the extent of advance of the chisel.However, the use of chiselling for routine work to achieve bell-outs is not encouragedbecause the control of chiselling is more difficult than the use of a mechanical under-reamer.

(13) Soft sediments. For sites with a deep layer of very soft sediments, sufficientadhesion may develop such that the casing may become stuck and may break at theconnections if excessive torque is applied during extraction.

(14) Piles in landfill and chemically contaminated ground. Bored pile constructionin landfill has potential problems associated with venting of methane gas, disposal ofcontaminated spoil, sudden loss of drilling fluids in voided ground and hazards ofunderground fire and surface explosion,

(15) Cavernous marble. The potential problems of pile construction in karstic groundinclude risk of necking at locations of weak superficial deposits, difficulty of seating on aninclined rock surface, the possible need to ream through thin slabs or treat weak materialsunderlying the slabs, potential loss of drilling fluid leading to bore instability, base heave,oozing in of soft cavity infill giving rise to sinkholes, excessive erosion of soil under highfluid pressure, etc. Expedients which may be adopted to assist pile construction in theseground conditions have been given in the literature (e.g. Chiu & Perumalswamy, 1987;Mitchell, 1985; Tan et al, 1985; Tang, 1986; Li, 1992). Where multiple cavities are found,telescopic permanent casings may be used but, if possible, one length of casing is preferredbecause it is comparatively more difficult to control the installation of multiple casings.

7.3.5 Potential Problems during Concreting

(1) General. The final concreted level should be at a sufficient distance above therequired trimmed level to allow removal of the surface laitance. The concreted level shouldpreferably be higher than the groundwater level to ensure concrete integrity. Where thetrimmed level is at depth and the concreted level is below the groundwater level, the problemof the water head exceeding the concrete head can be alleviated by partially filling the emptybore with granular material and topping up with water where a permanent liner is left in, orfilling the bore with spoil prior to extracting the temporary casing. If either bentonite slurryor water is added and mixed with the soil in the ground by the drilling equipment to assistwith the installation of the temporary casing (i.e. -'mudding-in'), the concreted level shouldbe coincident with the piling platform level.

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Regardless of the method of concrete placement, it is difficult to properly placeadditional concrete on top of the previous lift after the temporary casing has been withdrawn.

(2) Quality of concrete. A high-slump, self-compacting mix is necessary in order toensure that the concrete flows between the reinforcement bars and fills the entire cross sectionof the bore. Concrete with low workability is a major cause of defects. To minimizesegregation, honeycombing and bleeding resulting from high water content, the use of aplasticizer additive may be beneficial.

In bored pile construction, the radial effective stress in soil may be significantlyreduced, such as in the pile section bored under water and ahead of casing. For such cases,the concrete pressure plays a pivotal role in restoring the radial effective stress, and the slumpof concrete and the time during which concrete remains fluid will control the shaft resistancethat can be achieved.

For piles where concreting is carried out in an unlined bore free of water and withample room for free movement of aggregates between bars, a typical concrete slump of100 to 150 mm will generally be acceptable. Where concrete is placed by tremie , aminimum slump of about 150 mm / 175 mm should be adopted.

It would be advisable to check the slump of every concrete load. Flow table tests maybe a more appropriate method for assessing the flow properties and cohesiveness of a highworkability mix in tremie concrete. No extra water or other constituent materials should beallowed to be added to ready-mix concrete on or off site.

Concrete in pile shaft should not be vibrated. If this were done, there would be a riskof the vibrated concrete arching onto the side of the casing and being lifted during casingextraction. Reliance is therefore placed on the energy of the free-falling concrete to achieveself-compaction.

(3) Quality of grout. Grout constituents for mini-piles should be mixed thoroughlyto produce a consistent colloidal grout. In general, a high-speed mixer is preferred to a low-speed paddle type mixer.

A useful discussion on the design of a grout mix is given by Bruce & Yeung (1984).Strict quality control of the constituent materials and the grouting procedure is essentialbecause the effect of improper grouting will be accentuated by the small diameter of the piles.

The range of quality control tests include measurements of fluidity (or viscosity),strength, bleeding and free expansion. The requirements for the tests are given inGeospec 1 : Model Specification for Prestressed Ground Anchors (GCO, 1989). In addition,the density of the liquid grout may be checked with the use of a mud balance whereappropriate. The setting time should also be noted.

Guidance on the acceptable limits of grout property, such as cementitious content,bleeding, free expansion, strength and fluidity, are given in the General Specification forCivil Engineering Works (Hong Kong Government, 1992a).

The volume of grout injected should be determined using a calibrated flowmeter,

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preferably cross-checked by means of a stroke counter on the pumping equipment.

(4) Steel reinforcement. Careful thought needs to be given to avoid closely-spacedreinforcement which may impede the flow of concrete, leading to integrity problems. Itwould be advisable to use a smaller number of larger bars with a minimum spacing of at least100 mm.

Proper design and fabrication of cages is necessary to ensure that failure of hoopreinforcement does not occur as the concrete is being placed in the pile. The case of a cagebeing grossly distorted by the wet concrete is usually evidenced by downward movement ofthe projecting bars. Fleming et al (1992) suggested the possible use of welded steel bandsin lieu of the normal helical binding to help prevent twisting of the cage during concreting.

In the case of mini-piles where special reinforcement couplers are used, it would beprudent to stagger these such that the minimum spacing between couplers is about 200 mm.

(5) Placement of concrete in dry piles. Experience in Hong Kong indicates thatconcrete of exceptionally low strength of the order of 7 to 10 MPa can result if concreteplacement is not controlled properly. The concrete must be placed in such a manner as toprevent segregation. The 'free-fall1 method of placing concrete has been found to begenerally satisfactory for piles up to about 40 m length provided the concrete falls directlyonto the base without striking the reinforcement or the sides of the bore. The direction ofdischarge should be controlled with the use of a rigid delivery tube positioned centrally overthe pile. It is good practice to use a full-length delivery tube but experience suggests that theconcrete may be placed successfully with the use of a short length of delivery tube providedthat the concrete is not deflected or impeded during the fall. Where a full length tube is notused, an expedient sometimes adopted is to discharge cement grout down the bore prior toplacing the first load of concrete in order to coat the reinforcement to reduce the effects ofsegregation. For raking piles, a full-length delivery pipe should always be used to minimizethe risk of segregation.

The interior surface of any temporary casing must not have lumps of fines adheringto it as a result of penetration of cohesive strata, and this can be checked by visual inspection.The lumps are liable to be dislodged by the concrete and form inclusions.

Ideally, the concreting should be carried out in one continuous operation. In the casewhere concrete delivery is delayed, the concrete already placed may start to bleed or partiallyset and laitance may be formed. This will, lead to poor joints between successive lifts.

Where water has accumulated.at the base of the pile, there is a risk of the cementbeing leached out leading to weaker concrete (Pratt, 1986), Thorburn & Thorburn (1977)suggested that if the depth of water accumulating within the bore exceeds 50 mm between thetime of removal of the downhole pump and deposition of the first batch of concrete, the waterlevel should be permitted to reach equilibrium and a tremie pipe used for concreting.Expedients sometimes adopted such as depositing some dry cement prior to discharge ofconcrete should be .'discouraged. It is a fallacy to assume that the greater density of-concretewill resist the water, as the hydraulic balance will only operate whilst the concrete retains itsfluidity. The Hong Kong Institution of Engineers (1981) recommended that where the waterinflow rate exceeds 0.3 litres/second, the tremie method should be used for concreting. In

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certain cases, instead of waiting for the water level to reach steady-state, it may beworthwhile to consider filling the bore with water, as valuable time can be saved and the borewould suffer less from stress relief and disturbance under the seepage forces.

(6) Placement of concrete in piles constructed under water on bentonite. Concreteplacement in piles constructed under water on bentonite is invariably carried out using atremie and requires good workmanship and close supervision. Problems have been reportedin the literature (e.g. Humpheson et al, 1986) with inferior concrete at the base of piles wherethe concreting operation is not properly controlled. Care should be taken to ensure that theconcrete flows freely and continuously through the tremie pipe. The tremie pipe should bewatertight and of sufficient strength. It is important to maintain the discharge end of thetremie pipe below the upper surface of the rising concrete at all times. The tremie pipeshould preferably be placed at a depth of between 2 m to 3 m below the concrete surface.Surging (i.e. lifting and lowering) of the tremie pipe should be avoided.

In the case of barrettes, a sufficient number of tremie pipes should be used to ensurethat the surface of the concrete rises uniformly within the excavation to minimize the risk ofbentonite slurry being trapped.

A plug of vermiculite or other suitable material should be used as an initial separationlayer between the first batch of concrete and the water in the open-ended tremie pipe tominimise the risk of segregation.

If the tremie pipe is lifted too high off the pile bottom at the start of concreting, thesudden discharge of concrete could cause intermixing and segregation, resulting in a softbase. Fleming & Sliwinski (1977) suggested the initial lifting should be limited to 100 mm.

The concrete must retain sufficient workability for 'plug' flow to take place, i.e. thealready-placed concrete is displaced by the newly-placed concrete as a whole. If the concretepartially sets, the newly-placed concrete may tend to rise above the fold f concrete by flowingalong the side of the tremie pipe. In this case, the filter cake on the wall of the bore will notbe scoured effectively and the concrete may contain inclusions.

In the case where the concrete mix is of insufficient workability or there is a longdelay in concrete delivery, the tremie pipe could become blocked. The time lapse betweenbatching and placement of concrete should be minimized as far as practicable. If the tremiepipe is raised to clear the blockage and attempts are made to re-insert into the concrete tocontinue concreting, the pile will be certain to contain inclusions.

(7) Concrete placement in continuous-flight auger piles. In continuous-flight augerpiles, the skill of the operator is important during the concreting stage in ensuring pileintegrity. The rate of concrete or grout injection and the rate of extraction of the auger mustbe properly co-ordinated to avoid necking. Extensive instrumentation has been developedoverseas to help monitor the drilling and concreting operation (e.g. Derbyshire et al, 1989)but this is still uncommon in Hong Kong.

(8) Extraction of temporary casing. The temporary casing should be clean andsmooth and free from distortions that may affect pile integrity during casing removal. Thecasing must be extracted along the axis of the pile.

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The workability of concrete will reduce if the time taken for concreting is excessive.Premature stiffening of the concrete is also possible when there is water absorption into dryaggregates or when too finely-ground or recently-ground cement is used. If this occurs, thereis a risk that the partially set concrete is lifted or damaged as the casing is removed. Thecasing may have to be left in to avoid potential damage to the concrete. In this case, anassessment of potential loss of pile capacity that results from the unintentional leaving of thetemporary casing should be made.

Defects could arise if water-filled or slurry-filled cavities created during excavationexist outside the casing and the casing is extracted too rapidly with insufficient concrete head.In this case, as concrete flows to partially fill the cavities, a bulb with a neck on top mayresult if the water within the cavities cannot flow away rapidly (Figure 47). This problemwill be exacerbated if the concrete mix is of insufficient workability and may necessitate theuse of a permanent liner in stratum where such cavities are likely to form.

Where a permanent casing is required inside the temporary casing, care should betaken to ensure that concrete or debris does not become lodged between the two casings.Otherwise, the permanent casing could also be lifted. Depending on the nature of theoverburden materials, consideration should be given to backfilling the void between thepermanent casing and the soil with a suitable material. The permanent casing should haveadequate thickness to avoid possible bursting or collapse.

(9) Effect of groundwater. Where there are significant hydraulic gradients in highlypermeable ground (e.g. tidal conditions near a river or piling in the vicinity of groundwaterpumping), there is a risk of leaching of cement and washing out of aggregates in newly-placed concrete. Steep interfaces between permeable strata and cohesive soils along whichgroundwater flows under significant hydraulic head can also provide the conditions necessaryfor such attack (Thorburn & Thorburn, 1977). When groundwater leaching is deemed to bea potential problem, a permanent casing of sufficient length should be used.

A case history of necking resulting from the combined effect of an upward flow ofartesian water and the presence of loose sand is discussed by Hobbs (1957). Relief pipesattached to the reinforcement cage have been used successfully in projects elsewhere torelieve artesian water pressures during concreting.

An unusual case history concerning problems with rock-socketed piles inmudstone andsiltstone is reported by Stroud (1987). In this case, the relatively small amount of waterseepage during pile bore excavation was sufficient to work the mudstone spoil into a pastebut insufficient to wash it off the walls. The paste was subsequently plastered around thebore by the cleaning bucket and caused a substantial reduction in shaft resistance. Theremedial solution adopted was to replace the piles, taking due care to add water to the shaftto ensure washing action as the cleaning bucket was introduced.

(10) Problems in soft ground. Defects may arise when forming bored piles in verysoft ground with undrained shear strengths of less than about 15 to 20 kPa. The lateralpressure of the wet concrete could exceed the passive resistance of the soft soils and bulgeson the pile shaft may occur. On the other hand where the concrete head within the casingis insufficient and the concrete mix is of relatively low workability, there is a possibility ofthe formation of 'necked' shaft due to concrete arching across the casing or due to soil

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bulging into the concrete.

Near the head of the pile, the lateral pressure of the wet concrete may be low andfurther reductions are possible due to friction as the casing is extracted. Under suchcircumstances, it is possible for the very soft soil to squeeze into the pile section and causenecking. The risk of this happening may be overcome by a permanent casing.

(11) Cut-off levels. The concreted level should be such that when the concrete withlaitance is cut down to the cut-off (or trimmed) level, the concrete will be homogeneous andsound. Where the specified cut-off level is low and at depth below ground surface, it maybe difficult to achieve the least length of concrete to be trimmed consistent with minimizingwastage and the time involved in cutting down. In the case of concrete being placed underbentonite, the top portion of the concrete column may be particularly prone to intermixingwith the bentonite cake scoured off the side of the bore.

7.3.6 Potential Problems after Concreting

(1) Construction of adjacent piles. Relatively 'green' concrete may be damaged bydriving piles in close proximity or due to ground movements associated with excavations.

When adjacent bored piles are constructed close to a newly-concreted pile, there is arisk of 'pile connection', i.e. the relief of stresses upon bore excavation may be sufficient toallow the partially set concrete to flow laterally, particularly where there is soft ground.

Careful thought should be given to planning the sequence of pile construction.

(2) Impact by construction plant. Cases have been known where cracks are inducedin the piles due to impacts by construction plant. Piles are particularly vulnerable when thepiling platform level is subsequently reduced exposing the tops of the piles. Alternatively,piles can be cracked when the projecting reinforcement bars are hit, sometimes by the pilingrig itself or the service crane during moves. Close supervision is necessary to prevent impactby construction plant.

(3) Damage during trimming. Damage may be caused to the concrete when ill-considered means are adopted to trim the pile. This could give rise to disputes as to whetherit is the Main Contractor or the Piling Sub-Contractor who is responsible for the cracks.

Where mechanical-controlled means are used to trim the pile head, it is recommendedthat the last half a metre or so of the concrete should be trimmed by hand-held pneumatictools for better control to minimise the possibility of the pile column being damaged.

(4) Cracking of piles due to thermal effects and ground movement. Large-diameterpiles are liable to crack under thermal stresses. Where the pile is adequately reinforced, thecracks are likely to be distributed throughout the depth of the section and are generally of noconcern. However, problems of interpretation of integrity tests may arise as to whether thecracks are structurally significant.

Excavation of basements after pile installation will give rise to ground movement and

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hence tension forces in the piles. Where piles are not adequately reinforced, significanthorizontal cracks may occur, affecting the settlement characteristics of the piles. Pilesconstructed beneath basements prior to excavation should be provided with adequate fulllength reinforcement to take the potential tension loading that may be generated by theexcavation.

7.4 INSTALLATION OF HAND-DUG CAISSONS

7.4.1 General

The construction of hand-dug caissons has been described in detail by Mak (1993) andoutlined in Section 3.4.3.

Guidance notes on standard good practice on the construction of hand-dug caissons arepublished by the Hong Kong Institution of Engineers (1981). This document covers keyaspects of construction considerations as well as supervision and safety.

7.4.2 Assessment of Condition of Pile Base

(1) Caissons in saprolite. For caissons founded in saprolite, insitu tests that can becarried out to assess the condition of the founding material upon completion of excavationinclude plate bearing tests (Sweeney & Ho, 1982) and continuous penetration tests using aGCO probe (a lightweight probing test) (Evans et al, 1982). Ku et al (1985) suggested thatat least three penetration tests should be made in the base of each caisson to assess the degreeand depth of any softening.

In carrying out the GCO probing test, standard equipment and testing procedure asdetailed in Geoguide 2 : Guide to Site Investigation (GCO, 1987) should be adopted. Thetests should be undertaken to at least 1 m below the pile base and the results reported as thenumber of blows for each 100 mm penetration (designated as the GCO probe blowcount, Np).Evans et al (1982) suggested that Np is roughly equivalent to SPT N value. This approximatecorrelation enables an assessment of whether the base condition is consistent with the designassumptions.

Core drilling may be carried out through tubes cast into a pile with the use of a tripletube core barrel to assess the condition of the base interface. The coring is typically extendedto not less than 600 mm below the pile base. It is important that attention is given to the useof an adequate flushing medium and its proper control for success in retrieving the core.

(2) Hand-dug caissons in rock. The discussion given in Section 7.3.3 concerningmachine-dug piles founded in rock is also relevant to hand-dug caissons. Thomas (1984)suggested that closed circuit television inspection can be carried out to confirm the interfacecondition for hand-dug caissons.

For hand-dug caissons bearing on rock, the base should be inspected to examine ifthere are sub-vertical seams of weaker rock or weathered material. Where present, theseshould be excavated to sufficient depth below the bottom and the local excavation plugged

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with suitable grout or concrete, prior to commencement of concreting of the pile shaft.

7.4.3 Potential Installation Problems and Construction Control Measures

(1) General There are a number of case histories in Hong Kong involving the useof hand-dug caissons in unfavourable ground conditions. In these cases, the caissons wereabandoned part way through the contract and replaced with an alternative pile type (Mak etal, 1994).

Potential problems during concreting relate to the quality of the concrete and adequacyof the reinforcement cage, together with the procedure of concrete placement. Reference maybe made to Section 7.3.5.

(2) Problems with groundwater. The construction of a hand-dug caisson below thegroundwater table can induce piping failure (i.e. hydraulic base failure). In coastalreclamation sites where the groundwater table is high and soft or loose superficial depositsextend to considerable depths, excessive inflow and bore instability may occur, leading toground loss and settlement around the site (Mackey & Yamashita, 1967b), and possiblecasualties within the caissons. Sudden base failure, probably due to an excessive differentialhydraulic head between the outside and the inside of the excavation has also been observedin very dense granitic saprolite with average SPT N values of about 70 to 80 prior toconstruction.

It is often difficult to assess the porewater pressure distribution and seepage gradientsbecause of the heterogeneity of the weathering profile and possible presence of structuraldiscontinuities including relict joints, erosion pipes, fault and dykes. As reported by Mortonet al (1980), the measured differential heads between the inside and the outside of a caissoncan be between 10% and 97% higher than that estimated based on the assumption of anisotropic, homogeneous aquifer and a simplified flow pattern.

Heavy seepage flow into the bottom of a caisson may cause weakening of the soilthrough slaking, leaching and dispersion. Loosening (or possible damage of bonding betweensoil grains) of initially dense to very dense saprolite can take place under significantgroundwater flows, as observed by Has well & Umney (1978).

Dewatering during caisson construction can cause extensive groundwater drawdownresulting in excessive ground settlement and may result in damage to surrounding utilityservices and structures. Chan & Davies (1984) observed that the average settlement ofbuildings supported on piles founded in completely weathered granite is 2 to 3 mm for everymetre head of drawdown.

The water discharged from the pumps should be collected in a sedimentation tank andchecked regularly to determine the quantity of fines being removed. This would assist in theidentification of zones with excessive loss of fines and give an early warning of the possibilityof subsidence or collapse of caisson rings in that area. Such ground loss may also lead toexcessive settlement of the ground surface.

(3) Base heave and shaft instability. Excessive differential head or hydraulic gradient

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and unstable ground could lead to collapse of the excavated face, rapid inflow of mud andwater, and heaving of the caisson base. In extreme situations, voids can be created in theground adjacent to the caissons and can lead to formation of sinkholes if ground loss isexcessive.

The rate of base heave has been found to be variable between sites, and between pilesin any one site (Shirlaw, 1987). In some cases, heave occurs quickly and can only berecognized by counting the number of buckets of arisings for each working shift. Themechanism of base heave is generally thought to be related to slaking, swelling and softeningof the soils which are a function of the degree of weathering and can be promoted by stressrelief and high seepage gradient (Chan, 1987). Alternatively, the bonded structure of thesaprolite may collapse as the material starts to yield under low effective stresses leading toexcess pore water pressures and therefore softening in situations where the material is in ametastable state (Lam, 1990).

Some weathered granites have been observed to exhibit a pronounced tendency forswelling and loosening at low effective stresses (Stroud & Sweeney, 1977; Davies & Henkel,1980). Mackey & Yamashita (1967a) observed that the zone of loss of soil strength was asmuch as 9 m away from the caisson. A possible cause of significant base heave and shaftinstability could be improperly backfilled site investigation boreholes or the presence of oldwells.

If excavation has to proceed below the apparent rock surface where caisson rings willnot be constructed, the risk of caisson instability arising from the presence of weathered rocksoutside the unsupported shaft possibly under a high water head should be carefullyconsidered. Local grouting of the soil-rock interface may be necessary in order to minimisethis problem.

(4) Base softening. It is common for softening to occur rapidly in granitic saprolitein the base of excavations below the water table (Philcox, 1962; Mackey & Yamashita,1967a). The susceptibility to softening is related to the degree of weathering. Somecompletely weathered granites swell rapidly when the effective stress is reduced to a lowvalue (Davies & Henkel, 1980).

Evans et al (1982) observed significant softening of a caisson base down to a depth of0.8 m, about 70% of the shaft diameter. The degree of softening increased with the lengthof time between completion of excavation and commencement of concreting. It was furtherobserved that upon concreting, re-compression of the softened base took place to a depth ofabout 50% of the pile diameter over a period of 10 days. Grouting of the pile base wascarried out at a maximum pressure of 300 kPa but the re-compression of the softened materialwas not significant in this instance. If there are lengthy delays to the placement ofreinforcement and concrete, consideration may be given to constructing a concrete plug at thebottom of the pile in order to limit the effects of stress relief.

Endicott (1980) reported similar findings of base softening but found from load testson short length concrete plugs that the base stiffness was satisfactory, with the load resistedby skin friction. However, to improve confidence level and alleviate the concern of longterm behaviour of caissons with a soft base, the pile base was grouted to achieve a givenprobe test resistance.

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Even in the situation where the general groundwater table has been drawn down, somedisturbance to the shaft of the bore will be inevitable due to stress relief and possible seepagegradient built up around the pile. This is highlighted by the results of horizontal plate testsin completely decomposed granite reported by Whiteside (1986). In these tests, the disturbedzone appeared to be fully re-compressed at a stress level ranging from 400 to 500 kPa, andit is notable that this stress level is substantially in excess of the vertical effective stress andthe likely pressure of the wet concrete.

(5) Effects on skin friction. In difficult ground conditions, forepoling stakes may bedriven into the ground ahead of the excavation to provide temporary support prior to thecasting of concrete liner for each lift. These timber stakes are typically left in the ground andcould potentially result in reduced skin friction.

Where there is a tendency for high seepage gradients and base heave, the ground maybe subject to softening around the pile and hence result in reduction in skin resistance. If thebore is allowed to cave in, loosening of the surrounding ground will result. Tests to evaluatethe available frictional resistance of the caisson rings can be carried out from within caissonsusing a special jacking frame (Sweeney & Ho, 1982; Sayer & Leung, 1987).

(6) Effects of blasting. Where blasting is used to break up obstructions or expediteexcavation in rock, consideration should be given to assessing the effects on relatively greenand mature concrete in adjacent caissons, as well as on caisson ring stability where boreexcavation is not complete.

(7) Cavernous marble. Houghton & Wong (1990) discussed the potential problemsassociated with construction of hand-dug caissons in karstic ground conditions. The principalproblem is the need for dewatering during construction which could lead to sinkholeformation (Chan, 1994a). The use of hand-dug caissons in karstic marble is stronglydiscouraged.

(8) Safety and health hazards. The particular nature and procedure adopted in hand-dug caisson construction has rendered this operation one of the most accident-prone pilingactivities in Hong Kong. The most common causes of accidents include persons falling intothe excavation, falling objects, failure of lifting gear, electrocution, ingress of water/mudflow, concrete ring failure, and asphyxiation. Furthermore, the working environmentconstitutes significant health hazards arising principally from the inhalation of silica dust.

Concern for safety and health hazards must start at the design stage and continue untilcompletion of the works. Training courses for workers and their supervisors should bepromoted. General guidance aimed at site operatives is provided by the Hong KongInstitution of Engineers (1981).

(9) Construction control Precautionary measures which could be adopted tominimise the effects of groundwater drawdown and ground loss include the construction ofa groundwater cut-off (e.g. sheet piles or perimeter curtain grouting coupled with well pointsor deep wells) which encloses the site, the use of recharge wells in the aquifer undergoingdrawdown (Morton et al, 1981), and advance grouting at each caisson position prior toexcavation. Reference may be made to Shirlaw (1987) on the choice of grout for caissonconstruction. Care should be taken to control the grouting pressures to avoid excessive

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ground movement.

Where deep well dewatering is deemed to be unwarranted, the use of pressure reliefwells constructed prior to commencement of excavation may be considered to reduce the riskof high hydraulic gradients developing during construction. This is particularly relevantwhere there is a risk of artesian water pressure at depth.

The presence of old wells or underground stream courses will affect the effectivenessof the pre-grouting operation. In addition, where fractures are induced in the ground duringgrouting as a result of using an inappropriate grout type or lack of control of the groutingprocess, the permeability and hence the rate of softening may increase which could lead tobase heave.

An alternative means of control is phasing of caisson construction sequence in orderto limit ground movements and ground water drawdown. Where caissons are sunk on a groupbasis, one or two caissons may be advanced first to serve as deeper dewatering points for theother caissons.

Where poor ground is encountered, grouting may be carried out locally to helpstabilize the soil for further excavation. Alternatively a steel casing may be installed throughthe soft ground. Any voids resulting from over-excavation or caving should be backfilledwith concrete of similar quality as the lining.

Where significant base heave has been observed, the surrounding ground is likely tohave been disturbed and both the shaft friction and the base resistance may be affected. Acareful review of the design for the affected caissons will need to be made.

The design of the linings should be examined for suitability and may need to beexamined after construction, as for any other structural temporary works. In assessing theeffects of blasting on relatively 'green' concrete, reference may be made to Mostellor (1980)who suggested limiting ppv values of 6, 13 and 25 mm/sec for a concrete age of 12, 24 and48 hours respectively as a very rough guide.

In addition to ensuring strict compliance with safety requirements and implementationof precautionary measures, it is important that sufficient instrumentation comprisingpiezometric and movement monitoring of the adjacent ground and structures is included tocontrol the excavation operation. The monitoring results should be regularly reviewed toassess the need for remedial measures.

Possible early signs of instability should be taken seriously and investigatedthoroughly. Excessive excavation depths and hence the risk of base heave will be reducedif rational design methods are adopted to avoid overly-conservative pile designs.

7.5 INTEGRITY TESTING OF PILES

7.5.1 Role of Integrity Testing

The most direct tests of pile integrity and performance under load are physical coring

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and static load tests. Both methods have limitations. Static load tests are not very effectivein determining pile integrity (Section 7.5.3). Physical coring can provide samples for visualexamination and for compression testing. However, physical coring can only examine a smallportion of the cross-sectional area and usually cannot sample important areas such as areasoutside the reinforcement and hence, it can only provide a partial check. Non-destructiveintegrity testing has been used to augment these tests in assessing structural integrity of piles.Provided the limitations of integrity tests are understood and allowed for, these tests canprovide a useful engineering tool for quality control. Although the tests are intrinsicallyindirect, they are relevant as comparative tests and can act as a means of screening largenumbers of nominally similar piles. This allows a reasoned and logical approach in theselection of piles for further investigation or compliance tests.

The tests can generally be carried out rapidly and without causing significant disruptionto the works. They can be cost-effective in that defective works or inadequate proceduresmay be identified at an early stage of foundation construction. The test results can usuallybe displayed on site and a qualified operator can judge the validity of the data and recogniseany potential defects from a preliminary assessment.

As a large number of piles can be tested, integrity testing can play an important rolein encouraging higher construction standards and promoting self-imposed improvements ininstallation techniques and quality control.

7.5.2 Types of Non-destructive Integrity Testing

(1) General The most commonly-used types of integrity testing in Hong Konginclude sonic logging (sometimes referred to as sonic coring), vibration (sometimes referredto as impedance or transient dynamic response) tests, echo (or seismic or sonic integrity)tests, and dynamic load testing. It should be noted that integrity testing can be used to assessthe integrity of a concrete pile but cannot predict pile capacity.

The principles and limitations of these tests are briefly summarised in the following.Other types of integrity tests include radiometric and electrical methods and stress wave tests(Fleming et al, 1992) which have been suggested and used with limited success elsewhere buthave not yet been introduced in Hong Kong. Reference may be made to Weltman (1977) fora summary of the principles of these tests.

(2) Sonic logging. This test is based on acoustic principles and essentially measuresthe propagation time of sonic transmission between two piezo-electric probes placed in plastictubes, or more usually metal tubes, cast into a pile. In general, the concrete/tube couplingis better with metal tubes. Plastic tubes, if used, must be sufficiently robust under the headand temperature of the wet concrete and during the lifting of the reinforcement cage. Plastictubes have also been found to be more prone to erroneous readings. Alternatively, soniclogging can be performed inside cored holes within a pile.

The tubes (usually 40 to 50 mm in diameter) are filled with water to provide acousticcoupling for the transmission. Both the emitter and receiver probes are lowered to the baseof the tubes and raised by a hand winch calibrated for depth at a rate of about 200 mm/sec.With the transmission frequency of about 10 Hz, this corresponds to a sonic pulse every

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20 mm. Alternatively, metal wheels with a depth encoder can be used.

Each arriving signal is used to produce a variation in intensity of an oscilloscope scanand is modulated to a series of black-and-white lines. Alternatively, the output can be in theform of a printout consisting of a plot of pulse time against depth. Any increase inpropagation time or loss of signal, which are indicative of poor quality concrete or defects,can be easily detected by comparing the signals one above the other. The complete trace canbe recorded on a polaroid camera or the results can be stored digitally. The scale of any partof the display may be blown up to allow a detailed examination. The emitter and receiverprobes may be lifted up to different levels so as to better define the extent of the defects.This arrangement should be used to check for the presence of horizontal cracks.

As the recorded signal is, to a certain extent, a function of the sensitivity of the signalconditioning equipment and the pre-selection of the threshold strength of the arriving signal,standardisation of equipment is essential.

Guidance on the number of tubes to be employed for different pile sizes is given byTijou (1984). The positions of the emitter and receiver probes can be varied in the tests toimprove the accuracy in the identification of the extent of defects (Figure 48). Tests usinga single tube can also be carried out. In this case, the tube should be made of plastic insteadof steel because the latter is a better transmitter of acoustic energy than concrete, and henceit is liable to affect the acoustic paths and give false results about the integrity of the concrete.

The main objective of sonic logging is to check the homogeneity of the concrete.Sonic logging can detect the presence of defects including honeycombing and segregation,necking, presence of foreign material (i.e. inclusions) and cracks. However, it is not capableof identifying the nature of the defects. Moreover since the tubes are normally placed insidethe reinforcement cage, sonic logging is generally not capable of identifying problems withinadequate peripheral concrete cover to reinforcement.

Controlled laboratory and field tests have been reported by Stain & Williams (1991)in the assessment of the effects of various types and sizes of anomalies on sonic loggingresults, and the effect of signal 'skipping' round the anomaly via the access tubes.

Sonic logging is generally used in cast-in-place piles or barrettes. As the test relieson a cross-hole method, there is no depth limitation associated with signal damping problems.However, there is a limit on the maximum distance between tubes for a reliable sonic traceto be obtained. Where sonic logging is adopted on large-diameter bored piles or barrettes,an additional tube may need to be installed in the centre of the pile to improve the dataquality. This may cause difficulties with heavily reinforced pile sections as the presence oftubes could be a contributory factor in the formation of defects. Also, poor bonding betweenthe tube and the concrete may result in anomalous response. For sonic logging, a pre-selection of the piles to be tested has to be made.

(3) Vibration (impedance) test. These tests are based on the measurement of thedynamic response of piles in the frequency domain. In its original form, the test involves theuse of an electro-dynamic vibrator to impose a sinusoidal force of constant amplitudecontaining energy over a broad frequency band, preferably from 0 to 5 000 Hz. Adevelopment of this test is the transient dynamic response method in which the transient

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frequency response of the pile to a single blow is analysed using Fast Fourier Transformtechnique. In this method, a small hand-held hammer fitted with an internal load cell is usedin lieu of the vibrator, and a vibration transducer (preferably an accelerometer but it can bea geophone) determines the resulting velocity at the pile head. The hammer must be able togenerate an impulse of the above frequencies. The results and the method of interpretationare identical for both types of test.

For the tests, the pile head should be prepared by trimming to sound concrete, andsometimes a layer of cement mortar is cast over the pile head. Preparation of the pile headshould be done at least one day before the test if mortar is used. The test is normally carriedout at least four days after casting of the pile.

The results are presented in the form of a mobility diagram in which the mechanicaladmittance (pile head velocity, Vt, per unit applied force, Fp) is plotted against excitationfrequencies, f. A typical trace is shown in Figure 49. In principle, the physicalcharacteristics that can be derived from the results are :

(a) Dynamic pile head stiffness (Kd) - This is the slope of thelow frequency (i.e. < 100 Hz) linear portion of the graphfrom the origin to the first peak. This value is sensitive tothe stiffness of the pile shaft under compression.

(b) Condition of anchorage at pile toe - The position of the firstresonant frequency (or peak on the trace) depends on theend condition of the pile. For a pile toe that is rigidlyconstrained (end bearing pile), the first resonant frequencyis given by Vc/4Lres where Vc is the average wave velocityin concrete and Lres is the resonating length. For anunconstrained pile toe (friction pile), the first resonantfrequency is Vc/2Lres.

(c) Resonating length (Lres) - Resonant peaks at highfrequencies occur at frequency intervals of Vc/2Lres.

(d) Characteristic mobility (M0) - The average value of Vt/Fp

from the trace is termed the characteristic mobility. This isgiven by the expression M0 = l/(pc Vc Ac), where pc isthe concrete density and Ac is the concrete cross-sectionalarea. For a given force, piles with a smaller section willhave a greater mobility. Thus, the relative concrete quality(or conversely the cross-sectional area if the strength isknown) can be assessed.

(e) Damping factor (Dc) - Damping of the signal by theinteraction of soil and pile is described by the ratio of themobility, Vt/Fp, at resonance (peaks) to that at anti-resonance (troughs) on the trace. Hence the greater theamplitude of the sinusoidal wave form, the less thedamping.

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Vibration tests are suitable for identifying anomalies such as cracks, poor jointing andnecking of piles. A guide to the interpretation of the test results is given in Table 26.

Vibration testing, although based on sound theory, is not a precise analytical tool. Thelimitations of the test may be summarised as follows :

(a) The signal is easily damped for piles with a length todiameter ratio of about 20 in stiff and dense soils and 30 inloose soils. Resonant peaks may be difficult to identify inpractice. For tubular piles, closed circuit televisioninspection may provide an alternative means of assessingpile integrity where signal damping is excessive (Evans etal, 1987).

(b) The wave velocity in concrete, Vc, has to be assumed inorder to calculate the resonating length, L^s- If Lres isknown, the average value of Vc can be calculated. Theassessment will not identify small but perhaps structurallysignificant variations in Vc through weak concrete zones.

(c) Small but abrupt changes in pile cross section (e.g.transition from the cased to the uncased bore) can oftengenerate resonant behaviour that is not structurallysignificant. On the other hand, the test may not be sensitiveto gradual changes in pile section.

(d) The test is unable to quantify the vertical extent of sectionchanges or the lateral position of defects.

(e) The test may not be able to detect vertical cracks.

(f) Subjective errors are possible, particularly for piles withcomplex and multiple resonance. A range of digital signalprocessing techniques, including digital integration andsignal averaging, may be adopted to aid interpretation (Chanet al, 1987). These advanced techniques must be used withextreme caution to avoid spurious results.

Where the number of joints in a precast pile is small and the condition of the splicingis good, the presence of joints is not necessarily a limitation to the use of vibration tests.

It is possible to carry out a computer simulation of the pile geometry and groundcharacteristics in advance of site testing. This simulation may be useful in enabling theengineer to correlate a doubtful curve with the probable kind of irregularity.

(4) Echo (seismic or sonic integrity) tests. The principle of echo tests is based on thedetection of a reflected echo or longitudinal wave returning from some depth down the pile.The measured time of travel of the vibration wave together with an assumed propagationvelocity enable the acoustic length to be determined. The test is normally carried out at least

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seven days after casting of the concrete.

There are two generic time domain echo type tests, namely sonic echo and pulse echo.Reference may be made to Ellway (1987) and Reiding et al (1984) for a summary of theprinciples of operation and interpretation of the tests. Forde et al (1985) also described theimprovements in time domain analysis of echo traces through the use of an auto-correlationfunction to detect reflections in the velocity-time signal.

In the echo test, the pile is struck by a hammer and the resulting vibration signal (e.g.velocity) is measured at the pile head by means of a geophone or an accelerometer. Ingeneral, longer pulses are used to detect defects at greater depths whilst shorter pulses areused for possible defects at shallow depths. After digital filtering of extraneously low andhigh frequency oscillations, the signals can be range-amplified to magnify the response.Random noise can also be reduced by signal-averaging techniques. Identification of reflectiontime and determination of echo phase can be done using signal processing techniquesincluding auto-correlation and cross-correlation methods.

Examples of typical test results are given in Figure 50. The phase of the reflectedwave provides a means of discriminating reflections from large bulbs or severe necks (orcracks) which constitute fixed and free surfaces respectively.

The limitations of the test may be summarised as follows :

(a) It is suitable for bored piles and single length H-piles andprecast piles, but jointed piles or piles with jointed casingmay cause problems.

(b) Reflections from surfaces of intermediate stiffness such assmall bulbs or necks can cause frequency-dependent phasedistortions of the signal making interpretation more difficult.

(c) In the case of anomalies near the pile head, the response canbe distorted to such an extent as to give rise to problems ofsignal filtering.

(d) The penetration of the signal into the pile is limited by shaftfriction. A high shaft friction will reduce pile length thatcan be tested. Under normal circumstances, it is generallyunlikely that a reflection can be detected for a pile with alength to diameter ratio of greater than 30.

(e) Site vibrations (evg. from construction plant) could affect thesignal. This effect may be minimised by analysing repeatedhammer blows and by signal averaging.

(f) It is capable of identifying well-defined cracks, particularlynear the pile head. However, the signal is less clear fordiagonal cracks.

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(g) It is insensitive to changes in concrete quality as an averagesonic velocity for concrete has to be assumed in theinterpretation. Any inclusion needs to be significant enoughto cause a reflection of the signal and this depends more onits dynamic and acoustic properties than on its strength.

Both the echo tests and vibration tests involve excitation of the pile head andmeasurement of the dynamic response to vibration. In principle, a single signal of a hammerblow can be analysed both in the time and frequency domains. There is a recent trend toattempt to combine the results to produce a trace referred to as an impedance log, whichprovides a vertical section through the pile (Paquet, 1992). However, this should be treatedwith caution as the number of variables involved are such that the impedance log may not beunique and precise.

(5) Dynamic load test (DLT). Dynamic load tests are high strain tests whereby stresswaves are generated by the impact of the pile with a piling hammer. Apart from detectingdefects in piles, dynamic load tests can be used to predict pile capacity. In the tests,sufficient force should be delivered to the pile such that a minimum pile penetration of about2 to 3 mm/blow is achieved where practicable, particularly if it is required to provide aprediction of the pile capacity. The stress wave will be reflected from the pile toe and anyirregularities in the pile shaft. The hammer impact and wave reflections are monitored withthe use of strain gauges and accelerometers. Further details of the tests and its applicationin the prediction of pile capacity are given in Section 8.4.

The results from the instrumentation are expressed as time history plots of the forceand velocity. Rausche and Goble (1979) suggested the use of a damage classification factor,j8z, which is defined in terms of changes in impedance (equation 7.1) as follows :

.................... - (7.4)

where Z2 = pile impedance above a given level where there is a significant change inimpedance

Zi = pile impedance below the same given level

The tentative classification scheme proposed by Rausche & Goble (1979) is reproducedin Table 27. This simplified method is related to the extent of pile cross-section that is leftafter the damage, and is based on the tacit assumption that the soil resistance immediatelybelow the point of damage is negligible.

The limitation of this method of integrity testing is that small cracks tend to close upduring the hammer blow, and only major damage can be identified. The presence of smallcracks can be detected using the sonic logging tests.

Broms & Bredenberg (1982) showed that if the time required to close a crack and thereflected stress wave are measured, the width of the crack may be calculated. An importantdistinction between a crack and significant damage is that the latter will become worse whilea crack will diminish as driving becomes harder. Fleming et al (1992) suggested that a crackof about 1 mm width would be a lower bound of detection by dynamic pile testing.

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7.5.3 Practical Considerations in the Use of Integrity Tests

The choice of the appropriate type of integrity tests should be made in relation to thetype of pile, the ground conditions, and the anticipated construction defects. It is essentialto have a basic understanding of the principles of the tests and their limitations.

Integrity tests are generally indirect tests and therefore cannot definitively identifywhether the defects, if any, will significantly affect the pile behaviour under load. Thus, theresults alone cannot serve as the basis for a sound engineering decision on the acceptabilityor otherwise of the pile. In all cases, experienced interpretation is required and the resultsof the interpretation must be considered in conjunction with the pile construction records.

Prior to conducting integrity testing, it is prudent to plan the course of actions thatneed to be taken if anomalies are detected.

It should be noted that integrity tests cannot be used to predict pile capacity. Therunning of integrity tests is valuable in that the results which exhibit anomaly could be usedas the basis in selection of piles for loading tests, thus permitting a much better appreciationof the relative performance of the pile population.

Dynamic load tests are somewhat special in that the tests can be used as integrity testsand can predict pile capacity. However, dynamic load tests have not yet been accepted foracceptance tests, unless they are calibrated with the appropriate static load tests. The piledriving analyser (PDA) testing associates with dynamic load tests may be used for thefollowing proposes :

(a) to identify in conjunction with piling records, doubtful pilesfor investigation or static load tests,

(b) to check the consistency of hammer efficiency,

(c) to assess the structural integrity of a pile, and

(d) to check the adequacy of the final set criterion as derivedfrom a pile-driving formula.

Tijou (1984) reported typical correlations established in Hong Kong between dynamicand static pile head stiffness for various types of driven and bored piles, and betweenpropagation velocity from sonic logging and unconfined compressive strength of concrete.These correlations should however be treated with caution as the database may not besufficiently representative for firm conclusions to be drawn.

It is important that a proper specification is drawn up which should clearly state theperformance requirements of the tests, the parameters to be measured, the means ofinterpretation and how the results should be reported. If the test data are presented in astandardised way, the results can be easily compared and contrasted.

It is essential that careful thought be given to the planning of an integrity testingprogramme. The testing should be properly integrated into the works construction

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programme with suitable stop or hold points included to allow the results to be fullyassimilated, examined and interpreted. Time should also be allowed for the possible need foradditional testing or investigation to supplement the integrity tests. Normally, a minimumof five percent of piles in one project are subject to integrity tests.

It should be recognized that only an acoustic anomaly may be identified by integritytests and this may not necessarily correspond to a structural defect. Despite the fact thatcracks and other minor defects may not influence the load-settlement performance of a pilein the short term, the long term performance may be impaired as a result of corrosion ofreinforcement, spalling of concrete or reduction in effective concrete sections. The engineershould consider appropriate means of investigating possible anomalies identified by integritytests including exposing the pile sections where practicable.

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8. PILE LOAD TESTS

8.1 GENERAL

Given the many uncertainties inherent in the design and construction of piles, it isdifficult to predict with accuracy the performance of a pile. The best way to establish pilebehaviour is to carry out a load test.

Load tests can be carried out on preliminary piles to confirm the pile design or onworking piles as proof load tests.

There are two broad types of pile load tests, namely static and dynamic load tests.Static load tests are generally preferred because they have been traditionally used and alsobecause they are perceived to replicate the long term sustained load conditions. Dynamicload tests are usually carried out as a supplement to static load tests. The failure mechanismin dynamic load test may be different from that in static load test. A recent innovation is aquasi-static load test whereby a controlled explosion in a pressure chamber is used to generatethe test load and measurements are made with a load cell and a laser beam. However, suchtests have not been carried out in Hong Kong but reference may be made to Birmingham &Janes (1989), Janes et al (1991) and Middendorp et al (1992) for details of the testingtechnique and the method of interpretation. Lee et al (1993) described a 'simple pile loadtest' system for driven tubular piles which comprises a separable pile shoe and a reduced-sizesliding core for a rapid determination of the separate components of skin friction and end-bearing, however, the experience with this in Hong Kong is limited.

In this Chapter, the different types of load tests which are in common use aredescribed. Details of pile instrumentation and information that can be derived from theinstrumented load tests are given.

8.2 TIMING OF PILE TESTS

For cast-in-place piles, the timing of a load test is dictated by the strength of theconcrete or grout in the pile. Weltman (1980b) recommended that at the time of testing, theconcrete or grout should be a minimum of seven days old and have a strength of at leasttwice the maximum applied stress.

With driven piles, there may be a build-up of pore water pressure after driving butdata in Hong Kong are limited. Lam et al (1994) reported that for piles driven intoweathered meta-siltstone the excess pore water pressure built up during driving took only oneand a half days to dissipate completely.

Results of dynamic load tests reported by Ng (1989) for driven piles in loose graniticsaprolites (with SPT N values less than 30) indicated that the measured capacities increasedby 15% to 25% in the 24 hours after installation. The apparent 'set up' may have resultedfrom dissipation of positive excess pore water pressure generated during pile driving or there-establishment of radial stresses which could have been affected by pile whipping.

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As a general guideline, Weltman (1980b) recommended that a driven pile should betested at least three days after driving if it is driven into a granular material and at least fourweeks after driving into a clayey soil, unless sufficient local experience or results ofinstrumentation indicate that a shorter period would be adequate for dissipation of excess porepressure. It should be noted that increase of pile bearing capacity has also been found tooccur in granular soils during a period of several weeks.

8.3 STATIC LOAD TESTS

8.3.1 Reaction Arrangement

(1) General To ensure stability of the test assembly, careful consideration shouldbe given to the provision of a suitable reaction system. The geometry of the arrangementshould also aim to minimize interaction between the test pile, reaction system and referencebeam supports. It is advisable to have, say, a 10% to 20% margin on the capacity of thereaction against maximum test load.

(2) Compression tests. There are three types of reaction system which can beused : kentledge, tension piles and ground anchors.

Kentledge is commonly used in Hong Kong (Figure 51). This involves the use of deadweights supported by a deck of steel beams sitting on crib pads. The area of the crib shouldbe sufficient to avoid bearing failure or excessive settlement of the ground. It isrecommended that the crib pads are placed at least 1.3 m from the edge of the test pile tominimize interaction effects (Institution of Civil Engineers, 1988). If the separation distanceis less than 1.3m, the surcharge effect from the kentledge should be determined and allowedfor in the interpretation of the load test results.

Tension piles used to provide reaction for the applied load (Figure 52) should belocated as far as practicable from the test pile to minimise interaction effects. A minimumcentre-to-centre spacing of 2 m or three pile diameters, whichever is greater, between the testpile and tension piles is recommended. If the centre spacing between piles is less than threepile diameters, there may be significant pile interaction and the observed settlement of the testpile will be less than what should have been. If a spacing of less than three pile diametersis adopted, uplift of the tension piles should be monitored and corrections should be made forthe settlement of the test pile based on recognised methods considering pile interaction, suchas Poulos & Davis (1980).

To reduce interaction between the ground anchors and the test pile, the fixed lengthsof the anchors should be positioned a distance away from the centre of the test pile of at leastthree pile of diameters or 2 m, whichever is greater. Ground anchors may be used insteadof tension piles to provide load reaction. The main shortcomings with ground anchors arethe tendon flexibility and their vulnerability to lateral instability.

The provision of a minimum of four ground anchors is preferred for safetyconsiderations. Installation and testing of each ground anchor should be in accordance withthe recommendations as given in GCO (1989) for temporary anchors. The anchor loadshould be locked off at 110% design working load. The movements of the anchor should be

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monitored during the load tests to give prior warning of any imminent abrupt failure.

The use of ground anchors will generally be most suitable in testing a raking pilebecause the horizontal component of the jacking may not be satisfactorily restrained in otherreaction systems. They should be inclined along the same direction as the raking pile.

(3) Lateral load tests. In a lateral load test, two piles may be jacked against eachother (Figure 53). It is recommended that the centre spacing of the piles should preferablybe a minimum of ten pile diameters (Canadian Geotechnical Society, 1992).

Alternative reaction systems including a 'deadman' or weighted platform are alsoshown in Figure 53.

(4) Uplift tests. A typical arrangement for uplift tests is shown in Figure 54. Thearrangement involving jacking at the centre is preferred because an even load can be appliedto the test pile. The arrangement of applying load at one end of the beam is notrecommended because of risk of instability.

Reaction piles should be placed at least three test pile diameters, or a minimum of 2m, from the centre of the test pile. Where the spacing is less than this, corrections forpossible pile interaction should be made (Section 8.3.1(2)).

Osterberg (1989) described the use of a special load cell which can be installed at ornear the bottom of a pile to jack the pile section upwards whilst reacting against the base.

8.3.2 Equipment

(1) Measurement of load. A typical load application and measurement system consistsof hydraulic jacks, load measuring device, spherical seating and load bearing plates(Figure 51).

The jacks used for the test should preferably be large diameter low-pressure jacks witha travel of at least 15% of the pile diameter (or more if mini-piles are tested). A single jackis preferred where practicable. If more than one jack is used, then the pressure should beapplied using a motorized pumping unit instead of a hand pump. Pressure gauges should befitted to permit a check on the load. The complete jacking system including the hydrauliccylinder, valves, pump and pressure gauges should be calibrated as a single unit.

It is strongly recommended that an independent load-measuring device in the form ofa load cell, load column or pressure cell is used in a load test. The device should becalibrated before each series of tests to an accuracy of not less than 2% of the maximumapplied load (ASTM, 1995a).

It is good practice to use a spherical seating in between the load measuring device andbearing plates in a compression load test in order to minimise angular misalignment in thesystem and ensure that the load is applied coaxially to the test pile; Spherical seating ishowever only suitable for correcting relatively small angular misalignment of not more thanabout 3° (Weltman, 1980b).

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A load bearing plate should be firmly bedded onto the top of the pile (or the pile cap)orthogonal to the direction of applied load so as to spread the load evenly onto the pile.

(2) Measurement of pile head movement. Devices used for measuring pile headsettlement in a load test include dial gauges (graduated to 0.01 mm), linear variabledifferential transducers (LVDT) and optical levelling systems. A system consisting of a wire,mirror and scale is also used in lateral load tests.

In a compression or tension test, measurements should be taken by four dial gaugesevenly spaced along the perimeter of the pile to determine whether the pile head tiltssignificantly. The measuring points of the gauges should sit on the pile head or on bracketsmounted on the side of the pile with a glass slide or machined steel plate acting as a datumfor the stems. Care should be taken to ensure that the plates are perpendicular to the pile axisand that the dial gauge stems are in line with the axis.

In a lateral load test, dial gauges should be placed on the back of the pile with thestems in line with the load for measuring pile deflection (Figure 53). A separate systeminvolving the use of a wire, mirror and scale may be used as a check on the dial gauges. Thewire should be held under constant tension and supported from points at a distance not lessthan five pile diameters from the test pile and any part of the reaction system (StandardsAssociation of Australia, 1995). Rotational and transverse movement of the pile should alsobe measured.

LVDT can be used in place of dial gauges and readings can be taken remotely.However, they are susceptible to dirt and should be properly protected in a test.

The reference beams to which the dial gauges or LVDT are attached should be rigidand stable. A light lattice girder with high stiffness in the vertical direction is recommended.This is better than heavy steel sections of lower rigidity. To minimise disturbance to thereference beams, the supports should be firmly embedded in the ground away from theinfluence of the loading system (say 2 m from piles or 1 m from kentledge support). It isrecommended that the beam is clamped on one side of the support and free to slide on theother. Such an arrangement allows longitudinal movement of the beam caused by changesin temperature. The test assembly should be shaded from direct sunlight.

In an axial load test, levels of the test pile and reference beam supports should bemonitored by an optical levelling system throughout the test to check for gross errors in themeasurements. The optical levelling should be carried out at the maximum test load of eachloading cycle and when the pile is unloaded at the end of each cycle. The use of precisionlevelling equipment with an accuracy of at least 1 mm is preferred. The datum for the opticallevelling system should be stable and positioned sufficiently far away from the influence zoneof the test.

8.3.3 Test Procedures

(1) General Two types of load testing procedures are commonly used, namelymaintained-load (ML) and constant-rate-of-penetration (CRP) tests. The ML method isapplicable to compression, tension and lateral load tests, whereas the CRP method is used

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mainly in compression load tests.

The design working load (WL) of the pile should be pre-deteraiined where WL isdefined as the allowable load for a pile before allowing for factors such as negative skinfriction, group effects, redundancy, etc.

(2) Maintained-load tests. In an ML test, the load is applied in increments, eachbeing held until the rate of movement has reduced to an acceptably low value before the nextload increment is applied. It is usual practice to include a number of loading and unloadingcycles in a load test. Such cycles can be particularly useful in assessing the onset of plasticmovements by observing development of the residual (or plastic) movement with increase inload. Based on this information, Butler & Morton (1971) deduced critical load ratios forpiles in difficult geological formations. This concept can be used to assess the acceptancecriteria for load tests on contract piles as discussed by Cole & Patel (1992).

Different loading procedures have been proposed in the literature (e.g. Weltman,1980b; ASTM, 1995a; Standards Association of Australia, 1978). Loading procedurescommonly used in Hong Kong include those recommended in the General Specification forCivil Engineering Works (Hong Kong Government, 1992a) for government projects, by BOO(1990) for private developments and by Housing Authority (Yiu & Lam, 1990) for publichousing projects. Details of the loading procedures are summarised in Table 28.

For testing of piles which will be subject to negative skin friction, additional allowanceneeds to be made for the portion of pile shaft above the neutral point which acts in positivefriction during the load test.

When testing a preliminary pile, the pile should, where practicable, be loaded tofailure or at least to sufficient movement (say, a minimum of 5% of pile diameter). If thepile is loaded beyond 2 WL, a greater number of small load increments, of say 0.15 to 0.2WL as appropriate, may be used in order that the load-settlement behaviour can be betterdefined before pile failure.

In principle, the same loading procedures suggested for compression tests may be usedfor lateral and tension load tests.

(3) Constant-rate-of-penetration tests. The constant-rate-of-penetration test has theadvantage that it is rapid. However, the mobilised pile capacity may be influenced by strainrate effects, particularly in cohesive soils.

A constant strain rate of 0.25 to 1.25 mm/min and 0.75 to 2.5 mm/min is commonlyused for clays and granular soils respectively (ASTM, 1995a). The load should be suppliedby a hydraulic power pack and by regulating the rate of oil flow to the jack and monitoringthe pile movement with dial gauges. This procedure can control the rate of pile penetrationbetter.

Experience with the use of CRP tests in Hong Kong is limited." Tsui (1968) reportedthat two piles at the Ocean Terminal Building site which have been subjected to a maintained-load test followed by a CRP test showed similar capacities although the load-settlementcharacteristics are different. In general, CRP tests are less suitable for piles founded on rock

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or granular soils and can constitute a safety hazard if the increase in loading becomesexcessive. CRP tests are not suggested in Hong Kong given the ground conditions.

8.3.4 Instrumentation

(1) General. Information on the load transfer mechanism can be derived from a loadtest if the pile is instrumented. To ensure that appropriate and reliable results can beobtained, the pile instrumentation system should be compatible with the objectives of the test.Important aspects including selection, disposition and methods of installation should becarefully considered.

It is essential that sufficient redundancy is built in to allow for possible damage andmalfunctioning of instruments. Where possible, isolated measurements should be made usingmore than one type of equipment to permit cross-checking of results. An understanding ofthe proposed construction technique and a preliminary assessment of the probable behaviourof the pile will be helpful in designing the disposition of the instruments. Limitations andresolutions of the instruments should be understood.

(2) Axial load tests. Information that can be established from an instrumented axialload test includes the distribution of load and movement, development of skin friction andbase resistance with displacement. A typical instrumentation layout is given in Figure 55(a).

Strain gauges (electrical resistance and vibrating wire types) can be used to measurelocal strains which can be converted to stresses or loads. Vibrating wire strain gauges aregenerally preferred, particularly for long term monitoring, as the readings will not be affectedby changes in voltage over the length of cable used, earth leakage and temperature variation.

There are two types of vibrating wire strain gauges, namely surface mounting gaugesand embedment gauges for the measurement of steel and concrete strains respectively. Thesegauges generally have a maximum strain range of 3 000 microstrain (pee) and a sensitivity ofabout 1 /xe. Surface mounting gauges consist of a plucking coil, end blocks and a stem. Theend blocks are welded onto the pile body or reinforcement and the stem is fixed in betweenthe blocks. Embedment gauges consist of a plucking coil and a stem with a flange at eachend and are usually mounted between supports fixed to the pile or cast in concrete briquettesprior to mounting. With the latter method, the gauges are better protected but there is adanger that the concrete used for the briquette has a different consistency to that of the pile,giving rise to uncertainties when converting strains to stress. The use of strain gauges castin concrete briquettes is therefore liable to give unreliable results.

To measure axial loads, the strain gauge stems are orientated in line with the directionof the load (i.e. vertical gauges). One set of gauges should be placed near the top of the pileto facilitate calculation of the modulus of the composite section. Gauges should also beplaced close to the base of the pile (practically 0.5 m) with others positioned near stratumboundaries and at intermediate levels. A minimum of two and preferably three gauges shouldbe provided at each level where practicable.

For cast-in-place piles, provisions should be made to take a core through the pile shaftafter the load test. The concrete cores should be tested to determine the uniaxial compression

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strength, Young's modulus and Poisson's ratio. Bonded or unbonded sensing device, suchas electrical strain gauges or LVDT are recommended for measuring the Young's modulusand Poisson's ratio (ASTM, 1992). The Young's modulus of the composite section can beestablished from the moduli of concrete and steel reinforcement. This provides a means ofchecking the Young's modulus back-calculated from the strain gauges near the top of the pile.

If measurement of the development of normal stress at pile-soil interface is required,additional strain gauges can be orientated to have their stems perpendicular to the directionof load application (i.e. horizontal gauges), with one of their ends as close as possible to thepile-soil interface.

Other devices are available for measuring axial loads such as shaft load cells (Price& Wardle, 1983) and Mustran cells (Owens & Reese, 1982) but these are not commonly usedin Hong Kong.

The load cell developed by Price & Wardle (1983) as shown in Figure 55(b) may beused for measuring the load at pile base. The load transducer for the cell comprises a steeltube fitted with an internal vibrating wire gauge. Load is transferred to the transducer bysteel bars bonded into the concrete. Alternatively, a hydraulic load cell can also be used formeasuring the base load.

Rod extensometers which are mechanically operated can be used for measuring pileshaft movements at designated levels. The system consists of a PVC sleeve and analuminium or glass fibre rod with an anchor attached to its end. Monitoring the movementof the rod gives the corresponding pile shaft compression. It should be cautioned thatextensometers can easily get twisted or damaged during installation because of the slendernessof the rods. The risk of damage can be reduced by placing the rods on opposite sides of thepile, which offers better overall protection. Extensometers using standard steel pipes as thecasing, and steel bars alternating with ball bearings as the inner rods, are also not so easilydamaged.

An Osterberg load cell (Osterberg, 1989) may be used to separate the skin frictioncomponent from the end-bearing component. However, it should be noted that where thisis used to jack the pile shaft against the pile base, the skin friction mobilised will be in anopposite direction to that experienced by the pile shaft under compression loading.

In general, it is advisable to assess whether the results of the instruments correspondto the expected behaviour under the applied load at an early stage of the test. Anydiscrepancies noted during load application may be rectified and the test may be restartedwhere appropriate.

(3) Lateral load tests. The common types of internal instrumentation used in a lateralload test are inclinometers, strain gauges and electro-levels.

The deflected shape of a pile subject to lateral loading can be monitored using aninclinometer. The system consists of an access tube and a torpedo sensor. For cast-in-placepiles, the tube is Installed in the pile prior to concreting. For displacement piles such as H-piles, a slot can be reserved in the pile by welding on a steel channel or angle section priorto pile driving. The tube is grouted into the slot after driving. During the test, a torpedo is

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used to measure the slope, typically in 0.5 m gauge lengths, which can be converted todeflections. Care needs to be exercised in minimising any asymmetrical arrangement of thepile section or excessive bending of the pole during welding of the inclinometer protectivetubing. In extreme cases, the pile may become more prone to being driven off verticalbecause of these factors.

Strain gauges with their stems orientated in line with the pile axis can be used formeasuring direct stresses and hence bending stresses in the pile. They can also be orientatedhorizontally to measure lateral stresses supplemented by earth pressure cells.

Electro-levels measure changes in slope based on the inclination of an electrolytic fluidthat can move freely relative to three electrodes inside a sealed glass tube (Price & Wardle,1983; Hong Kong Institution of Engineers, 1995). The changes in slope can be convertedto deflections by multiplying the tangent of the change in inclination by the gauge length.The devices are mounted in an inclinometer tube cast into the pile and can be replaced if theymalfunction after installation. The use of electro-levels in piles has not been common inHong Kong.,

Earth pressure cells can also be used to measure the changes in normal stresses actingon the pile during loading. It is important that these pressure cells are properly calibratedfor cell action factors, etc. to ensure sensible results are being obtained.

8.3.5 Interpretation of Test Results

(1) General. A considerable amount of information can be derived from a pile loadtest, particularly with an instrumented pile. In the interpretation of test results for design,it will be necessary to consider any alterations to the site conditions, such as fill placement,excavation or de watering, which can significantly affect the insitu stress level, and hence thepile capacity, after the load test.

(2) Evaluation of failure load. Typical load-settlement curves together with somepossible modes of failure are shown in Figure 56. Problems such as presence of a soft claylayer, defects in the pile shaft and poor construction techniques may be deduced from thecurves where a pile has been tested to failure.

It is difficult to define the failure load of a pile when it has not been loaded to failure.A commonly-used definition of failure load is taken to be that at which settlement continuesto increase without further increase in load; alternatively, it is customarily taken as the loadcausing a settlement of 10% of pile diameter. In practice, the failure or ultimate loadrepresents no more than a bench mark such that the safe design working load can bedetermined by applying a suitable factor of safety. In the case where ultimate failure has notbeen reached in a load test, a limiting load may be defined which corresponds to a limitingsettlement or rate of settlement.,•

An estimate of the ultimate or failure load may also be made by hyperbolic curve-fitting as proposed by Chin (1970). However, such a procedure can be inherently unreliableeven if the extrapolation is carried out to a movement of only 10% pile diameter, especiallywhere a pile has not been tested to exhibit sufficient plastic movement. In addition, it also

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has drawbacks as it does not deal with the base resistance and shaft friction load separatelynor does it take into account of elastic shortening, (Fleming, 1992). The danger associatedwith gross extrapolation is highlighted by the results of load tests reported by Yiu & Lam(1990). Notwithstanding the above, the method proposed by Chin (1978) may be useful inthe diagnosis of whether a pile has suffered structural damage during a load test.

Methods have been proposed in the literature for separating the skin friction and baseresistance components from the load-settlement relationship at pile head (e.g. Van Wheele,1957; Hobbs & Healy, 1979). These methods are approximate and may not be appropriatefor long slender piles or in complex and variable ground conditions. Fleming (1992),proposed a new method for single pile settlement prediction and analysis based on animprovement on the use of hyperbolic functions, However, the experience of using thisprediction method in Hong Kong is still very limited.

(3) Acceptance criteria. From the load-settlement curve, a check of pile acceptabilityin terms of compliance with specified criteria can be made. In Hong Kong, three sets ofacceptance criterion are generally used for compression load tests (Table 28) :

(a) 90% criterion proposed by Brinch Hansen (1963) adopted inthe General Specification for Civil Engineering Works(Hong Kong Government, 1992a) and mainly used forpublic developments,

(b) criteria proposed by Davisson (1972), which have beenmodified and adopted by the Buildings Ordinance Office(BOO, 1990) and mainly used for private developments, and

(c) criteria proposed by the Canadian Geotechnical Society(1992) and adopted by the Housing Authority (Yiu & Lam,1990).

There are other sets of acceptance criterion as adopted by local authorities, such asArchitectural Services Department (ASD). However, these are used only as internaldocuments.

It should be noted that the first term of the expressions in criteria (b) and (c) representsthe elastic compression of the pile shaft in an end-bearing pile (ie. with no skin friction).Where the pile is long and resists most of the applied load by skin friction, the actual elasticcompression will be less. The expressions are therefore not rational for friction piles and willresult in a less stringent requirement for acceptance. However, the expressions as they standgenerally lead to conservative estimates of failure load compared to other definitions ofultimate loads (Figure 57) and a further adjustment to reduce the maximum allowablesettlement for skin friction piles is likely to make it overly conservative. This practice istherefore not recommended.

Upon removal of the maximum test load, the permanent settlement of a long pile canbe 'locked in1 as a result of reversal of skin friction upon removal of the test load. Limitingresidual settlement criteria such as (b) and (c) above are therefore inappropriate for frictionpiles (Fraser & Ng, 1990) and unnecessarily conservative design may result in this case. In

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principle, a designer should concentrate on the limiting deflection at working load as well asthe factor of safety against failure or sudden gross movements. The limiting settlement ofa test pile at working load should be determined on an individual basis taking into accountthe sensitivity of the structure, the elastic compression component, effects of pile groupinteraction under working condition, and expected behaviour of piles as observed in similarprecedents.

In a tension test, reference may be made to Kulhawy & Hirany (1989) for a generaldiscussion of the background considerations. The use of Brinch Hansen's (1963) criterionmay not be suitable for tension piles which may fail abruptly in the absence of an end-bearingcomponent. A modified form of Davisson's (1972) criteria was suggested as follows(Kulhawy & Hirany, 1989):

= Elastic extension + 4 mm (8.1)

A slightly different expression, where the second term is 2.5 mm instead of 4 mm,was used by Davie et al (1993). The determination of the elastic extension is subject touncertainties associated with the load distribution down the pile, progressive cracking of theconcrete or grout, etc. It is suggested that the above criterion may be adopted, where theelastic extension is taken to be given by the initial linear portion of the load-extension curve.Based on the observations of uplift load test results of bored piles, Kulhawy and Hirany(1989) proposed to use the load corresponding to a pile head displacement of 13 mm as theuplift capacity of pile.

Different factors of safety may be appropriate when different definitions of failure loadare used. It would be rational to unify the definition of ultimate loads to permit comparisonand extrapolation of test results. In view of the shortcomings inherent in criteria (b) and (c)as discussed above, it is recommended that the acceptance criterion given in the GeneralSpecification for Civil Engineering Works (Figure 58) for the definition of failure load shouldbe adopted consistently, unless there is a valid reason for doing otherwise.

(4) Axial load tests on instrumented piles. The profile of shaft movement along a pileas determined by extensometers allows the shaft compression between any two points in thepile to be calculated from which the load distribution can be deduced (Tomlinson, 1994).

The load distribution down a pile can also be determined by strain gauges. From this,the mobilisation of skin friction and base resistance can be assessed.

The existence of residual stresses prior to application of test load, particularly fordriven piles, should be considered when the instrumentation results are back-analysed inderiving 'fundamental1 soil parameters. Significant residual stresses will affect the profile ofload distribution with depth and the apparent stiffness of the pile under compression ortension loading (Poulos, 1987). Altaee - e t - a l - (1992a & b) highlighted the importance ofmaking proper allowance for residual stresses in the interpretation of an instrumented piledriven into sand.

(5) Lateral load test. No performance criteria have been specified in the Building(Construction) Regulations (Hong Kong Government, 1993) and the General Specification forCivil Engineering Works (Hong Kong Government, 1992a) for piles under lateral loading.

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The limiting criteria on displacement and/or rotation have to be assessed by designers forindividual cases, taking into account factors such as sensitivity of structures, nature ofloading, etc.

The profiles of deflection, slope, bending moment, shear force and soil reaction areinterrelated and may be represented by differential equations. For instance, the profile of piledeflection and soil resistance may be deduced from the bending moment profile by doubledifferentiation and double integration respectively, allowing for the effect of bending stiffness.In practice, however, the accuracy of the measurements can have a profound influence on theparameters derived by this method and the results should be treated with caution.

Hirany & Kulhawy (1989) advocated a new approach for evaluating lateral load testresults. This consists of determining the variation of the apparent depth of rotation, definedas the ratio of the lateral displacement to the tangent of the slope of the upper part of thedeflected pile, with the applied load (Figure 59). This method can only be used if both thedisplacement and rotation of the pile top have been recorded. The variation in the apparentdepth of rotation will give a hint on the mode of failure, i.e. structural failure, rigid rotationof the shaft, yielding of soil in front, or yielding of soil behind the pile with a 'kick-out' ofthe tip (Figure 59).

(6) Other aspects of load test interpretation. Care should be taken in ensuring thatthe test load is maintained for a sufficient period since redistribution of load down the pileshaft may take place as observed by Promboon et al (1972). Premchitt et al (1988) alsoreported an increase of up to 10% in axial strains at points along the pile as time dependentload transfer moving progressively downwards took place when the test load was maintainedfor three days.

Endicott (1980) presented results of load tests carried out on caissons founded ingranitic saprolites at different times after construction. A significant increase in stiffness wasobserved after a six month delay which may be related to a recovery of strength of the soilwith time; however, the results may have been affected to a certain extent by the previousloading/unloading cycles.

Based on the findings of Tomlinson & Holt (1953), Malone (1990) cautioned about thepotential discrepancies in the building settlement and the rate of settlement as observed in apile test.

8.4 DYNAMIC LOAD TESTS

8.4.1 General

Various techniques for dynamic load tests are now available. These tests are relativelycheap and quick to carry out compared with static load tests. Information that can beobtained from a dynamic load test includes :

(a) static load capacity of the pile,

(b) energy delivered by the pile driving hammer to the pile,

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(c) maximum driving compressive stresses (tensile stress shouldbe omitted), and

(d) location and extent of structural damage.

8.4.2 Test Methods

Dynamic load test is generally carried out by redriving a prefabricated pile after acertain set-up period or by applying impact loading on a cast-in-place pile by a drop hammer.A standard procedure of carrying out a dynamic load test is given in ASTM (1995b).

The apparatus required for carrying out a dynamic pile load test includes a drivinghammer, strain transducers and accelerometers, together with appropriate data recording,processing and measuring equipment.

The hammer should have a capacity large enough to cause sufficient pile movementsuch that the resistance of the pile can be fully mobilised. A guide tube assembly to ensurethat the force is applied axially on the pile should be used.

The strain transducers contain resistance foil gauges in a full bridge arrangement. Theaccelerometers consist of a quartz crystal which produces a voltage linearly proportional tothe acceleration. A pair of strain transducers and accelerometers are fixed to opposite sidesof the pile, either by drilling and bolting directly to the pile or by welding mounting blocks,and positioned at least two diameter or twice the length of the longest side of the pile sectionbelow the pile head to ensure a reasonably uniform stress field at the measuring elevation.It should be noted that change of cross-section of the pile due to connection may affect theproportionality of the signals and hence the quality of the data. An electronic theodolite mayalso be used to record the displacements of the pile head during driving (Stain & Davis,1989).

In the test, the strain and acceleration measured at the pile head for each blow arerecorded. The signals from the instruments are transmitted to a data recording, reducing anddisplaying device to determine the variation of force and velocity with time.

8.4.3 Methods of Interpretation

(1) General. Two general types of analysis based on wave propagation theory,namely direct and indirect methods, are available. Direct methods of analysis apply tomeasurements obtained directly from a (single) blow, whilst indirect methods of analysis arebased on signal matching carried out on results obtained from one or several blows.

Examples of direct methods of analysis include CASE, IMPEDANCE and TNOmethod, and indirect methods include CAPWAP, TNOWAVE and SIMBAT. CASE andCAPWAP analyses are used mainly for displacement piles, although in principle they can alsobe applied to cast-in-place piles. SIMBAT has been developed primarily for cast-in-placepiles, but it is equally applicable to displacement piles.

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In a typical analysis of dynamic load test, the penetration resistance is assumed to becomprised of two parts, namely a static component, Rs, and a dynamic component, R^Three methods of analysis that are commonly used in Hong Kong are described below.

(2) CASE method. This method assumes that the resistance of the soil is concentratedat the pile toe. In the analysis, the dynamic component is given by,

Rd = J c Z V b ................... (8.2)

where jc = the CASE damping coefficientZ = impedance = Ep Ap /c

Ap = cross sectional area of the pileEp = Young's modulus of the pile

c = wave speed through the pileVb = velocity of pile tip

The appropriate jc is dependent on the type of soil at the pile toe and the actual piledimensions. A range of jc values appropriate to different soil types was proposed by Rauscheet al (1985) as given in Table 29. It should be noted that the recommended values of jc arepredominantly for end-bearing piles. In practice, jc values can vary significantly, particularlyin layered and complex ground conditions, causing potential errors in pile capacity prediction.

(3) CAPWAP method. In a CAPWAP (CASE Pile Wave Analysis Program) analysis,the soil is represented by a series of elasto-plastic springs in parallel with a linear dashpotsimilar to that proposed by Smith (1962). The soil can also be modelled as a continuumwhen the pile is relatively short. Using the measured acceleration-time data as the inputboundary condition, the program computes a force versus time curve which is compared withthe recorded data. If there is a mismatch, the soil model is adjusted. This iterativeprocedure is repeated until a satisfactory match is achieved between the computed andmeasured force-time diagrams.

The dynamic component of penetration resistance is given by,

R d = J s V p R s . . . ......... ---- . . . (8.3)

where js = Smith damping coefficientVp = velocity of pile at each segment

Input parameters for the analysis include pile dimensions and properties, soil modelparameters including the static pile capacity, Smith damping coefficient, js and soil quake (ie.the amount of elastic deformation before yielding starts), and the signals measured in thefield. The output will be in the form of distribution of static unit skin friction against depthand base response, together with the static load-settlement relationship up to about 1.5 timesthe working load. It should be noted that the analysis does not model the onset of pile failurecorrectly and care should be exercised when predicting deflections at loads close to theultimate pile capacity.

Results of CAPWAP analysis also provide a check of the CASE method assumptionssince the ultimate load calculated from the CAPWAP analysis can be used to calculate the

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CASE damping coefficient.

Sound engineering judgement is required in determining whether a satisfactory matchhas been achieved and whether the corresponding combination of variables is realistic.

(4) SIMBAT method. SIMBAT is developed mainly for cast-in-place piles. Thismethod is different from the other methods in that in addition to strain transducers andaccelerometers, an electronic theodolite is used for monitoring the temporary and permanentpile head movement during driving.

In the SIMBAT analysis,

Rd = Rs f(Vb) (8.4)

where f(Vb) denotes a function of the velocity of the pile tip.

An alternative formulation was suggested by Hansen and Denver (1980) for piledriving analysis as follows :

Rd = ' Z ( V 0 - 0 . 5 VO (8.5)

where V0 = first peak in velocity after the falling mass contacts the pile topV} = second peak in velocity upon arrival of the reflected wave at the pile topZ = Impedance (see Equation 8.2)

In this method, the soil is represented by a series of springs and dashpots (Stain &Davis, 1989). A series of impacts is applied to the pile using a drop hammer with the dropheight being progressively increased and decreased. The method of analysis is the same asin CAPWAP except that the displacement record obtained by the theodolite is used to verifyand correct the velocity data derived from the first integral of the acceleration data. Theupward and downward forces for each hammer blow are separated and the dynamic soilresistance for each blow is calculated. Experience with the use of this method in Hong Kongis, as yet, limited.

(5) Other methods of analysis. There are other methods of analysis such as thatproposed by Simons & Randolph (1985) and Lee et al (1988). These are generally based oninput of conventional soil mechanics parameters such as Young's modulus and density anddo not rely on empirical constants (i.e. damping factors and soil quake) as used in the aboveformulations. Experience with the use of these recent methods for practical problems ishowever limited.

8.4.4 Recommendations on the Use of Dynamic Load Tests

The reliability of the prediction of dynamic load test methods is dependent on theadequacy of the wave equation model and the premise that a unique solution exists when thebest fit is obtained within the limitation of the assumption of an elasto/rigid plastic soilbehaviour (Rausche et al, 1985), In addition, there are uncertainties with the modelling ofeffects of residual driving stresses in the wave equation formulation.

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A more detailed formulation of the wave equations such as CAPWAP or SIMBAT arepreferred for analysing dynamic load tests. These can be applied to non-homogeneous soilsor piles with a varying cross-sectional area. The static load-settlement response of a pile canalso be predicted. In practice, the wave equation analysis may be used to calibrate thedamping coefficients in a CASE analysis to permit more piles to be tested by the lessexpensive CASE method. This is a reasonable approach in a relatively uniform soil, providedthe CASE analyses are also calibrated against site-specific static load tests. As the field datacollected for a CASE analysis will be sufficient for a detailed wave equation analysis, thelatter should be carried out when the CASE analysis results are in doubt. In complex groundconditions, it is preferable to undertake more detailed wave equation analyses.

The results of dynamic load test may be used for design provided an adequate siteinvestigation has been carried out and the method has been calibrated against static load testson the same type of pile of similar length, cross section and under comparable soil conditions.It should be noted that a proper calibration of the analysis model has to be based on tests onpiles loaded to failure. A static load test which is carried out to a proof load is aninconclusive result for assessing the ultimate resistance of the pile.

Dynamic load tests may be used as an indicator of the consistency of the piles and todetect weak piles. By building up a reliable local database of piles tested statically anddynamically, dynamic load tests may be useful with confidence for determining pile capacitywithout static load tests, particularly for piles in area where access is difficult or on smallsites, or when time is limited.

00vo

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202

203

TABLES

to

2

205

LIST OF TABLES

Table PageNo. No.

1 Principal Rock and Soil Types in Hong Kong 207

2 Advantages and Disadvantages of Displacement Piles 209

3 Advantages and Disadvantages of Replacement Piles 210

4 Minimum Global Factors of Safety for Piles in Soil 211

5 Minimum Mobilisation Factors for Skin Friction and Base 212Resistance

6 Typical Values of Angle of Friction at Pile/Soil 213Interface in Granular Soils

7 Typical Values of Coefficient of Horizontal Pressure on 213Pile under Compression

8 Typical Values of Skin Friction Coefficient in Sand 214

9 Presumed Safe Vertical Bearing Pressure for Foundations 215on Horizontal Ground in Hong Kong

10 Classification of Marble 216

11 Shape and Rigidity Factors for Calculating Settlements 217of Points on Loaded Areas at the Surface of an ElasticHalf-space

12 Correlations between Drained Young's Modulus and SPT N 218Values for Weathered Granites in Hong Kong

13 Typical Values of Coefficient of Horizontal Subgrade 219Reaction

14 Positional Tolerances of Installed Piles 220

15 Calculation of Deflection of a Laterally-loaded Pile Group 220Using the Subgrade Reaction Method

16 Typical Efficiency of Pile Hammers 221

.17 Guidance on Minimum Weight of Pile Hammer 221

18 Possible Defects in Displacement Piles Caused by Driving 222

206

Table PageNo. No.

19 Defects in Displacement Piles Caused by Ground Heave 223and Possible Mitigation

20 Problems with Displacement Piles Caused by Lateral Ground 224Movement and Possible Mitigation

21 Problems with Driven Cast-in-place Piles Caused by 225Groundwater and Possible Mitigation

22 Typical Noise Levels for Different Piling Methods 226

23 Limits on Properties of Bentonite Slurry Suggested by 227Hong Kong Government (1992a) and Federation ofPiling Specialist (1977)

24 Example of a Specification on Bentonite Slurry Properties 228Adapted by Local Contractors

25 Causes and Mitigation of Possible Defects in Replacement 229Piles

26 Interpretation of Vibration Tests on Piles 233

27 Classification of Pile Damage by Dynamic Load Test Results 233

28 Loading Schedules and Acceptance Criteria for 234Pile Load Tests in Hong Kong

29 Suggested Values of CASE Damping Coefficients for 236Different Soils

Table 1 - Principal Rock and Soil Types in Hong Kong (Sheet 1 of 2)

Geological Period Age (Ma) Group Formation Principal Rock or Soil Types

SEDIMENTARY AND VOLCANIC ROCKS

QuaternaryHolocenePleistocenePleistocene - HolocenePleistocene - Holocene

Tertiary

Upper Cretaceous

Lower Cretaceous

Upper Jurassic - LowerCretaceous

<0.01<0.15

65

97.5

144

7130

135

140142

160

1701 / v/

Repulse BayVolcanic Group

Tsuen WanVolcanic Group

Hang HauSham WatChek Lap Kok(unnamed)

Ping Chau

KatOPort IslandPat Sin Leng

Ngo Mei ChauLong HarbourLai Chi ChongHigh IslandTai Mong TsaiPak TarnSai KungClear Water BayMang Kung UkSilverstrandMt DavisAp Lei ChauLantau

Sai Lau KungTai Mo ShanShing MunYim Tin TsaiTuen MunTsing Shan

Marine mud, marine sandMarine mud, marine sandAlluvial and estuarine clay, silt, sand & gravelSlope deposits

Siltstone

Conglomerate with coarse sandstoneConglomerate with coarse sandstoneConglomerate, sandstone & siltstone

Fine ash vitric tuffCoarse ash crystal tuffFine ash tuff with tuffiteFine ash tuffTuffiteRhyolite lavaCoarse ash tuffTrachydacite and rhyolite lavaTuffaceous mudstone, siltstone and brecciaEutaxiteCoarse ash crystal tuffFine ash vitric tuffFine ash vitric tuff and lava

Dacite lava with tuff and siltstoneCoarse ash crystal tuffFine to coarse ash tuffCoarse ash crystal tuffAndesite with tuff and tuffiteSandstone, siltstone and mudstone

Table 1 - Principal Rock and Soil Types in Hong Kong (Sheet 2 of 2)

Geological Period Age (Ma) Group Formation Principal Rock or Soil Types

Lower Jurassic

Permian

Carboniferous

Devonian

San TinGroup

Tolo Channel Formation

Tolo Harbour Formation

Lok Ma ChauYuen LongMa On Shan

Bluff Head Formation

Meta-sandstone and meta-conglomerateMarbleMarble

Sandstone and conglomerate

MAJOR INTRUSIVE IGNEOUS ROCKS

Upper Jurassic -Lower Cretaceous 136

136

138

147

148

152155

Lion RockSuite

Lamma Suite

Kwun Tong GraniteMt Butler GranitePo Toi GraniteKowloon GraniteSouth Lamma GraniteChi Ma Wan GraniteNeedle Hill GraniteD'Aguilar SyeniteMa Wan GraniteSha Tin Granite

Tsing Shan GraniteTai Lam GraniteSung Kong GraniteTai Po Granodiorite

Fine-grained monzograniteFine- and medium-grained monzograniteMedium- and coarse-grained monzograniteMedium-grained monzograniteMedium-grained monzograniteMedium-grained monzograniteFine- and medium-grained monzograniteQuartz syeniteFine- and medium-grained monzograniteMedium- and coarse-grained monzogranite

Medium- and coarse-grained monzograniteFine- and medium-grained monzograniteMedium- and coarse-grained hornblende-biotite monzograniteFine- and medium-grained granodiorite

Table 2 - Advantages and Disadvantages of Displacement Piles

Types of Pile Advantages Disadvantages

Large-displacementPiles

(a) Material of preformed section can beinspected before driving.

(b) Steel piles and driven cast-in-placeconcrete piles are adaptable to variabledriving lengths,

(c) Installation is generally unaffected bygroundwater condition.

(d) Soil disposal is not necessary.(e) Driving records may be correlated with

insitu tests or borehole data.(f) Displacement piles tend to compact

granular soils thereby improving bearingcapacity and stiffness.

(g) Pile projection above ground level and thewater level is useful for marine structuresand obviates the need to cast insitucolumns above the piles.

(h) Driven cast-in-place piles are associatedwith low material cost,

(a) Pile section may be damaged during driving.(b) Founding soil cannot be inspected to confirm the ground

conditions as interpreted from the ground investigationdata.

(c) Ground displacement may cause movement of, or damageto, adjacent piles, structures, slopes or utility installations.

(d) Noise may prove unacceptable in a built-up environment.(e) Vibration may prove unacceptable due to presence of

sensitive structures, utility installations or machinerynearby.

(f) Pile cannot be easily driven in sites with restrictedheadroom.

(g) Excess pore water pressure may develop during driving,resulting in negative skin friction on piles upon dissipation.

(h) Length of precast concrete piles may be constrained bytransportation or size of casting yard,

(i) Heavy piling plant may require extensive site preparation toconstruct a suitable piling platform in sites with poorground conditions.

(j) Underground obstructions cannot be coped with easily,(k) For driven cast-in-place piles, the fresh concrete is exposed

to various types of potential damage, such as necking,ground intrusions due to displaced soil and possible damagedue to driving of adjacent piles.

toovo

Small-displacementPiles

(a) As (a), (b), (c), (d), (e) and (g) for large-displacement piles.

(b) Cause less ground disturbance and lessvibration.

(a) As (a), (b), (d), (f) and (i) for large-displacement piles.

Table 3 - Advantages and Disadvantages of Replacement Piles

Type of Pile Advantages Disadvantages

Machine-dug Piles (a) No risk of ground heave induced by piledriving.

(b) Length can be readily varied.(c) Spoil can be inspected and compared with

site investigation data.(d) Structural capacity is not dependent on

handling or driving conditions.(e) Can be installed with less noise and

vibration compared to displacement piles.(f) Can be installed to great depths.

(a) Risk of loosening of sandy or gravelly soils during pileexcavation, reducing bearing capacity and causing groundloss and hence settlement.

(b) Susceptible to waisting or necking during concreting inunstable ground.

(c) Quality of concrete cannot be inspected after completionexcept by coring.

(d) Unset concrete may be damaged by significant water flow.(e) Excavated material requires disposal, the cost of which will

be high if it is contaminated.

Hand-dug Caissons (a) As (a) to (e) for machine-dug piles.(b) Use of comparatively economical labour

force.(c) Versatile construction method requiring

minimal site preparation and access.(d) Removal of obstructions or boulders is, relatively easy through the use of

pneumatic drills or, in some cases,explosives.

(e) Generally conducive to simultaneousexcavation by different gangs of workers.

(f) Not susceptible to programme delayarising from machine down time.

(g) Can be constructed to large diameters.

(a) As (a), (c) and (e) for machine-dug piles.(b) Hazardous working conditions for workers and the

construction method has a poor safety record.(c) Liable to base heave or piping during excavation,

particularly where the groundwater table is high.(d) Possible adverse effects of dewatering on adjoining land

and structures.(e) Health hazards to workers, as reflected by a high incidence

rate of pneumoconiosis and damage to hearing of caissonworkers.

toh-*o

211

Table 4 - Minimum Global Factors of Safety for Piles in Soil

Method of DeterminingPile Capacity

Minimum Global Factor of Safetyagainst Shear Failure of the Ground

Compression Tension Lateral

Theoretical or semi-empirical method notverified by load tests onpreliminary piles

3.0 3.5 3.0

Theoretical or semi-empirical method verifiedby a sufficient number ofload tests on preliminarypiles

2.0 2.5 2.0

Notes : (1) Table based on American Society of Civil Engineers (1993).(2) Assessment of the number of preliminary piles to be load tested is

discussed in Section 5.10.(3) Factor of safety against overstressing of pile materials should be in

accordance with relevant structural design codes. Alternatively,prescribed allowable structural stresses may be adopted as appropriate.

(4) In most instances, working load will be governed by consideration oflimiting pile movement, and higher factors of safety (or 'serviceability1

factors) may be required.

212

Table 5 - Minimum Mobilisation Factors for Skin Friction and Base Resistance

Material

Granular Soils

Clays

Mobilisation Factor for SkinFriction, fs

1.5

1.2

Mobilisation Factor for BaseResistance, fb

3-5

3-5

Notes : (1) Mobilisation factor for base resistance depends very much onconstruction. Recommended minimum factors assume goodworkmanship without presence of debris, etc. giving rise to a 'soft' toeand are based on available local instrumented load tests on friction pilesin granitic saprolites. Mobilization factors for base resistance alsodepend on the ratio of skin friction to base resistance. The higher theratio, the lower is the mobilization factor.

(2) Noting that the movements required to mobilise the ultimate baseresistance are about 2% to 5% of the pile diameter for driven piles, andabout 10% to 20% of the pile diameter for bored piles, lowermobilization factor may be used for driven piles.

(3) In stiff clays, it is common to limit the peak average skin friction to100 kPa and the mobilised base pressure at working load to a nominalvalue of 550 to 600 kPa for settlement considerations, unless highervalues can be justified by load tests.

(4) Where designer judges that significant mobilisation of base resistancecannot be relied on at working load due to possible effects ofconstruction, a design approach which is sometimes advocated (e.g.Toh et al, 1989; Broms & Chang, 1990) is to ignore the base resistancealtogether in determining the design working load with a suitablemobilisation factor on skin friction alone (e.g. 1.5). Base resistance istreated as an added safety margin against ultimate failure andconsidered in checking for the factor of safety against ultimate failure.

(5) Lower mobilisation factor for base resistance may be adopted for end-bearing piles provided it can be justified by settlement analyses that thedesign limiting settlement can be satisfied.

213

Table 6 - Typical Values of Angle of Friction at Pile/Soil Interface in Granular Soils

Pile/Soil Interface Condition dJ<t>'

Steel/sand

Cast-in-place concrete/sand

Precast concrete/sand

0.5 to 0.9

1.0

0.8 to 1.0

Legend

Angle of shearing resistance Angle of friction atpile/soil interface

Notes : (1) Table based on Kulhawy (1984).(2) Where steel is considered rough (corrugated),

upper end of the above range.' ratio will be near

Table 7 - Typical Values of Coefficient of Horizontal Pressure on Pile under Compression

Pile Type

Large-displacement piles

Small-displacement piles

Replacement piles

KS/K0

1 to 2

0.75 to 1.25

0.7 to 1.0

Legend :

Ks

Coefficient of earth pressure at restCoefficient of horizontal pressure on pile

Notes : (1) Table based on Kulhawy (1984).(2) In granular soils, K0 may generally be taken as (1 - sin $') where <£' is

the angle of shearing resistance unless they have been over-consolidated or preloaded.

(3) In clays, K0 will be related to the overconsolidation ratio, OCR(Meyerhof, 1988).

214

Table 8 - Typical Values of Skin Friction Coefficient in Sand

Type of Piles

Driven piles

Driven piles

Bored piles

Bored piles

Type of Soils

Loose sand

Dense sand

Loose sand

Dense sand

Shaft Friction Coefficient, 0

0.4 to 1.0

1.0 to 2.0

0.15 to 0.3

0.25 to 0.6

Notes : (1) Table based on Davies & Chan (1981).(2) Limited available load test results (Appendix A) indicate that at relative

settlement of 1 % pile diameter, j8 for saprolite are closer to the lowerbound values quoted in this table.

(3) Limited data from instrumented piles in Hong Kong suggest that thelimiting average shaft friction would be around 140 kPa for graniticsaprolites.

215

Table 9 - Presumed Safe Vertical Bearing Pressure for Foundations on HorizontalGround in Hong Kong

Category Rock (granites and volcanics)

PresumedBearingPressure

(kPa)

Ka)

Fresh to slightly decomposed strong rock of materialweathering grade II or better, with a total core recovery ofmore than 95% of the grade and minimum uniaxialcompressive strength of rock material (erc) not less than 50MPa (equivalent point load index strength, PLI50, not less than2MPa)

7500

Slightly to moderately decomposed moderately strong rock ofmaterial weathering grade II or III or better with a total corerecovery of more than 85 % of the grade and minimum ac notless than 25 MPa (PLI50 not less than 1 MPa)

5000

Kc)Moderately decomposed, moderately strong to moderatelyweak rock of material weathering grade III or IV or better,with a total core recovery of more than 50% of the grade

3000

Notes : (1) In using the table, the ground investigation should be planned, carriedout and supervised to standards recommended in GCO (1987).

(2) Terminology of rock descriptions recommended in GCO (1988) shouldbe followed. Rock material decomposition grades should be supportedby appropriate index tests, e.g. Schmidt hammer test, slake test, handpenetrometer, etc.

(3) Bearing values are for foundations with negligible lateral loads atfounding level.

(4) Self weight of the length of pile embedded in soil or rock does not needto be included in the calculation of bearing stresses.

(5) Minimum socket depth to be 0.5 m for category l(a) and 0.3 m forcategories l(b) and (c).

(6) Total Core Recovery is the percentage ratio of rock recovered (whethersolid intact with no full diameter, or non-intact) to the length of 1.5 mcore run in a drillhole and should be proved to a depth of 5 m belowthe base or three times the average width of the pile whichever is thegreater.

(7) Point load index strength of rock quoted is the equivalent value for50 mm diameter cores (i.e. PLI50 values).

(8) Use of presumptive safe bearing pressures does not preclude the needfor consideration of settlement of the structure.

216

Table 10 - Classification of Marble

MarbleClass

I

II

III

IV

V

MQDRange (%)

75.1 - 100

50.1 - 75

25.1 - 50

10.1 - 25

0 - 10

Rock MassQuality

Very Good

Good

Fair

Poor

Very Poor

Features

Rock with widely spaced fractures andunaffected by dissolution

Rock slightly affected by dissolution, orslightly fractured rock essentiallyunaffected by dissolution

Fractured rock or rock moderatelyaffected by dissolution

Very fractured rock or rock seriouslyaffected by dissolution

Rock similar to Class IV mable except thatcavities can be very large and continuous

Notes : (1) Table based on Chan (1994b).(2) In this system, Class I and Class II rock masses are considered to be a

good bearing stratum for foundation purposes, and Class IV and ClassV rock masses are generally unsuitable.

(3) Class III rock mass is of marginal rock quality. At one extreme, theClass III rating may purely be the result of close joint spacings inwhich case the rock may be able to withstand the usual range ofimposed stresses. At the other extreme, the Class III rating may be theresult of moderately large cavities in a widely-jointed rock. Thesignificance of Class III rock mass would need to be considered inrelation to the quality of adjacent sections and its proximity to theproposed foundations.

217

Table 11 - Shape and Rigidity Factors for Calculating Settlements of Points onLoaded Areas at the Surface of an Elastic Half-space

Shape

CircleCircle(rigid)SquareSquare(rigid)Rectangle:length/width

1.5235

10100

100010000

Shape and Rigidity Factor, Cs

Centre

1.000.791.120.99

1.361.521.782.102.534.005.476.90

Corner

0.640.790.560.99

0.670.760.881.051.262.002.753.50

Middle ofShort Side

0.640.790.760.99

0.890.981.111.271.492.202.943.70

Middle ofLong Side

0.640.790.760.99

0.971.121.351.682.123.605.036.50

Average

0.850.790.950.99

1.151.301.521.832.253.705.156.60

Note : Table based on Perloff (1975).

Table 12 - Correlations between Drained Young's Modulus and SPT N Values for Weathered Granites in Hong Kong

Drained Young's Modulusof Weathered Granites

(MPa)

1 . 8 N - 3 N

0.6 N - 1.9N

0.4 N - 0.8 N0.55 N - 0 . 8 N< 1.05 N

2 N - 2 . 5 N

1 N - 1.4 N

0.6 N - 1 N

0.2 N -0.3 N

3 N

0 . 6 N - 1 . 9 N(average 1.2 N)

L I N1.4 N1.7N

0.8 N1.6 N at depth

I N -1 .2N

1 N

0.7 N- 1 N

Basis

Pile load tests

Pile load tests

Pile load tests

Pile load tests

Pile load tests

Pile and plate load tests

Plate tests at bottom of hand-dug caissons

Horizontal plate tests in hand-dug caissons (unload-reload cycle)

Horizontal plate tests in hand-dug caissons (initialloading)

Multiple well pumping test and back analysis ofretaining wall deflection

Back analysis of retaining wall deflection

Settlement monitoring of buildings on pile foundations

Settlement monitoring of buildings on pile foundations

Back analysis of settlement of Bank of China Building

Range of SPTN Values

37 - >200

12-65

50 - 100100 - 150

> 150

25 - 160

50 - 100

50 - 200

35 - 250

47 - 100

47 - 100

25-5050 - 7575 - 150

up to 170

N/A

50 - 100

50 - 75

Reference

Fraser & Lai (1982)

Evans et al (1982)

Holt et al (1982)

Lamet al (1994)

Leung (1988)

Chan & Davies (1984)

Sweeney & Ho (1982)

Whiteside (1986)

Whiteside (1986)

Davies (1987)

Humpheson et al (1986, 1987)

Ku et al (1985)

Leung (1988)

Chan & Davies (1984)

toH-*00

219

Table 13 - Typical Values of Coefficient of Horizontal Subgrade Reaction

Consistency

n\i for dry or moist sand(MN/m3)

% for submerged sand(MN/m3)

Loose(N value 4-10)

2.2

1.3

Medium Dense(N value 11-30)

6.6

4.4

Dense(N value 31-50)

17.6

10.7

Notes : (1) The above % values are based on Terzaghi (1955) and are valid forstresses up to about half the ultimate bearing capacity with allowancemade for long-term movements.

(2) For sands, Elson (1984) suggested that Terzaghi f s values should beused as a lower limit and the following relationship as the upper limit :

nh = 0.19Dr116 (MN/m3)

where Dr is the relative density of sand in percent. Dr can be related toSPT N values and effective overburden pressure, (see Figure 6 ofGeoguide 1 : Guide to Retaining Wall Design (GEO, 1993)). Theabove equation is intended for sands and should be used with cautionfor saprolites. If this equation is used as a first approximation, it wouldbe prudent to determine the design value of Dr involving the use ofinsitu and laboratory density tests. In critical cases where the design islikely to be dominated by the behaviour under lateral loading, it isadvisable to carry out full scale load tests in view of the designuncertainties.

(3) Limited available load test results on piles in saprolitic soils in HongKong suggest that the % values can be bracketed by therecommendations by Terzaghi and the above equation by Elson.

(4) Other observed values of nh, which include an allowance for long-termmovement, are as follows (Tomlinson, 1994) :Soft normally-consolidated clays : 350 to 700 kN/m3

Soft organic silts : 150 kN/m3

For sands, % may be related to the drained horizontal Young modulus(5)(Eh') in MPa as follows :

' to .l.

where z is depth below ground surface in metres.(6) It should be noted that empirical relationships developed for

transported soils between N value and relative density are notgenerally valid for weathered rocks. Corestones, for example, cangive misleading high values that are unrepresentative of the soil mass.

220

Table 14 - Positional Tolerances of Installed Piles

Description

Deviation from specified position in plan,measured at cut-off level

Deviation from vertical

Deviation of raking piles from specified batterDeviation from specified cut-off level

Tolerance

Land piles

75 mm

1 in 75

Marine piles

150mm

1 in 25

1 in 2525 mm

Note : Table taken from General Specification for Civil Engineering Works (HongKong Government, 1992a).

Table 15 - Calculation of Deflection of a Laterally-loaded Pile Group Using theSubgrade Reaction Method

Principle Method Remarks

Reduction of soilmodulus

Reduce the % value in the direction of loading by areduction factor Rn for pile centre spacing of 8D, 6D,4D and 3D, where Rn = LOO, 0.70, 0.40 and 0.25respectively. Pile spacing normal to direction of loadhas no influence, provided it is greater than 2.5D.

Recommended byNAVFAC (1982) andCanadian GeotechnicalSociety (1992)

Reduction of soilresistance mobilizedat a given deflectionor increase indeflection at a givensoil resistance

Use a multiplier (fm or ym) to modify the H~<5h curvefor a single pile to obtain an effective H-<5h curve forthe pile group

Recommended byBrown et al (1988)based on results offield tests

Notes : (1) First method is widely used in design practice and is generally an adequateapproximation.

(2) Second method may be adopted where there are sufficient data from load tests.

221

Table 16 - Typical Efficiency of Pile Hammers

Type of Hammer

Drop hammer

Steam or air/pneumatic hammer

Diesel hammer

Hydraulic hammer

Typical Hammer Efficiencies (%)

70 - 80

60-70

35 - 70

75 -90

Notes : (1) Hammer efficiency corresponds to the ratio of actual energy transferredto the pile to the rated capacity of the hammer. The ratio is sometimesreferred to as the transfer ratio.

(2) Actual amount of energy transferred to the pile is best determined by-dynamic pile testing.

(3) The above are based on general experience in Hong Kong.

Table 17 - Guidance on Minimum Weight of Pile Hammer

Pile Type

Precast reinforced concretepiles

Precast prestressed concretepiles

Steel piles

Guidance

Weight of hammer about half the weight of the pile.

Generally, at least 3 tonnes, but 2 tonnes can be usedfor piles with a maximum length of 10 m andmaximum load of 45 tonnes. 4 tonnes should be usedfor long piles in compact materials. Drop should belimited to 300 to 400 mm in soft or loose soils and to300 mm in compact granular soils.

Varies with the capacity of steel pile. A minimum of2 tonnes for a pile with 60 tonne capacity.

Note : Table based on Healy & Weltman (1980) and Fleming et al (1992).

222

Table 18 - Possible Defects in Displacement Piles Caused by Driving

Pile Type

Steel piles

Concretepiles

Problems

Damaged pile top (head)(e.g. buckling, longitudinalcracking, distortion)

Damaged pile shaft(e.g. twisting, crumpling,bending)

Collapse of tubular piles

Damaged pile toe(e.g. buckling, crumpling)

Base plate rising relative to thecasing, loss of plugs or shoes incased piles

Damaged pile head(e.g. shattering, cracking,spalling of concrete)

Damaged pile shaft(e.g. fracture, cracking, spallingof concrete)

Damaged pile toe(e.g. collapsing, cracking,spalling of concrete)

Possible Causes

(a) Unsuitable hammer weight(b) Incorrect use of dollies, helmets,

packing(c) Rough cutting of pile ends(d) Overdriving

(a) Unsuitable hammer weight(b) Inadequate directional control of

driving(c) Overdriving(d) Obstructions

(a) Insufficient thickness

(a) Overdriving(b) Obstructions(c) Difficulty in toeing into rock

(a) Poor welding(b) Overdriving(c) Incorrect use of concrete plugs

(a) Unsuitable reinforcement details(b) Insufficient reinforcement(c) Poor quality concrete(d) Inadequate concrete cover(e) Unsuitable hammer weight(f) Incorrect use of dollies, helmets,

packing(g) Overdriving

(a) Excessive restraint on piles duringdriving

(b) Unsuitable hammer weight(c) Poor quality concrete(d) Inadequate or incorrect concrete

cover(e) Obstructions(f) Overdriving(g) Incorrect distribution of driving

stresses from use of incorrectdollies, helmets, or packing

(a) Overdriving(b) Poor quality concrete(c) Insufficient reinforcement(d) Inadequate or incorrect concrete

cover(e) Obstructions(f) Absence rock shoe where required

223

Table 19 - Defects in Displacement Piles Caused by Ground Heaveand Possible Mitigation

Problems

Uplift causing squeezing,necking or cracking of adriven cast-in-place pile

Uplift resulting in loss ofbearing capacity

Ground heave lifting pilebodily

Ground heave resulting inseparation of pile segmentsor units or extra tensileforces on the joints

Remedial Measures

None

Redrive piles

May not be necessaryfor friction piles

May be gently tappedor redriven.

Precautionary Measures

(a) Provide adequatereinforcement

(b) Plan driving sequence(c) Avoid driving at close

centres(d) Pre-bore(e) Monitor ground

movements

(a) Plan driving sequence(b) Allow for redriving(c) Avoid driving at close

centres(d) Pre-bore(e) Drive tubes before

concreting for driven cast-in-place piles

(f) Monitor pile movements

(a) Use small displacementpiles

(a) Plan driving sequence(b) Allow for redriving(c) Avoid driving at close

centres(d) Pre-bore(e) Consider other piling

systems

224

Table 20 - Problems with Displacement Piles Caused by Lateral Ground Movementand Possible Mitigation

Problems

Squeezing or waisting ofpiles or soil inclusionforced into a drivencast-in-place pile

Shearing of piles or bendsin joints

Collapse of casing prior toconcreting

Movement and damage toneighbouring structures

Remedial Measures

None

None

None, but if damage isminor, the pile may becompleted and used,subject to satisfactoryload test

Repair the structure.Change to a small-displacement orreplacement pilingsystem

Precautionary Measures

(a) Avoid driving at closecentres

(b) Allow concrete to set beforedriving nearby

(c) Pre-bore

(a) Plan the driving sequence(b) Avoid driving at close

centres(c) Pre-bore(d) Monitor pile movements

(a) Avoid driving at closecentres

(b) Pre-bore(c) Ensure that casing is thick

enough

(a) Plan the driving sequence(b) Isolate the structure from

driving(c) Use small-displacement piles(d) Pre-bore

225

Table 21 - Problems with Driven Cast-in-place Piles Caused by Groundwaterand Possible Mitigation

Problems

Water ingressduring drivingcasing andsubsequentdifficulties inconcreting

Bulging of pileand associatedwaisting above

Water ingress,causing softeningof the base (thismay becomeapparent onconcreting theshaft when thereinforcementmoves down thepile, possiblydisappearing fromthe pile head)

Causes

Loss of shoe or baseplate during driving

Failure of welds orjoints of tube

Failure of seal onjoints

Cracking of casingsections because ofincorrect distributionof driving stresses

Soft groundconditions(undrained shearstrength <1 5 kPa).Displacement ofground underhydrostatic head ofconcrete

Water-bearing sandsand gravels

RemedialMeasures

Replug withconcrete anddrive on

None

None

None

None

May be necessaryto redrive anotherpile through thefaulty pile

Precautionary Measures

(a) Use of gasket on shoeto exclude waterduring driving

(b) Use of pressure cap toexclude water

(a) Check integrity ofwelds prior to driving

(b) Take care in drivingto avoid hammerclipping any jointrings

(a) Good supervision toensure the joints areformed properly

(a) Care in driving anduse of correct packing

(a) Use of a pile typeemploying apermanent liner

(a) Good supervision isessential

(b) Check for wateringress by leaving thehammer resting on thebase before concretingthe shaft. If there iswater ingress, thiswill be apparent whenthe hammer is lifted

Note : Table based on Healy & Weltman (1980).

Table 22 - Typical Noise Levels for Different Piling Methods

PilingEquipment

Piles

LW(A)(continuousoperation)in dB(A) °B-1

L, xat 10 m nindB(A) an.,

Diesel Hammer

SteelPile

130121.7

11011

2.0

PrestressedConcrete

Piles

1264

(2.1)

1064

(1.9)

Double-acting

PneumaticHammer

SteelPile

1329

3.2

1089

3.4

Single-acting

PneumaticHammer

Steel Pile

1302

1092

—

Single-actingSteam

Hammer

SteelPile

1281

1091

—

Drop Hammer

SteelPile

123113.3

1099

3.3

PrecastConcrete

Pile

1142

982—

Vibrator

SteelPile/SteelSheet

1203

(3.2)

962

— •

InternalDrop

Hammer

SteelTubular

Pile

1124

(1.1)

914

(1.3)

Auger

BoredPile

1123

(2.4)

872

™ —

Grab, Chisel,Clamshell, Reverse

Circulation Drill

Large-diameterBored Pile, Barrette,

Diaphragm Wall

1117

4.9

876

3.9

Legend :

LW(A) A-weighted noise power levelLj The noise level that is exceeded for 1 % of the time interval consideredx Mean noise level in dB(A)n Number of observations

an.! Standard deviation

Notes : (1) Table based on Kwan (1985).(2) Figures in parentheses are for reference only as the sample size is too small.

toto

227

Table 23 - Limits on Properties of Bentonite Slurry Suggested by Hong KongGovernment (1992a) and Federation of Piling Specialists (1977)

Bentonite Propertyat 20°C

Density as suppliedto excavation

Density at base ofexcavation beforeplacing concrete

Viscosity

Shear strength (10minute gel strength)

pH value

Method of Testing

Mud density balance

Mud density balance

Marsh cone method

Fann viscometer

Shearometer

Fann viscometer

pH indicator paperstrips or electricalpH meter

Test Results

Hong KongGovernment(1992a)

< l. lOg/ml

< 1.25 g/ml

30 to 50 seconds

< 0.02 Pa.s(i.e. <20 cP)

1.4 to 10N/m2

4 to 40 N/m2

8 to 12

Federation ofPiling Specialists(1977)

< 1.10 g/ml

< 1.25 g/ml

30 to 90 seconds

< 20 cP

1.4 to 10 N/m2

4 to 40 N/m2

9.5 to 12

Table 24 - Example of a Specification on Bentonite Slurry Properties Adopted by Local Contractors

Bentonite Properties

Viscosity(sec)

Density(g/ml)

Fluid Loss(ml in 30 min)

'Cake' Thickness(mm)

pH

Sand Content %(% by Volume)

Shear Strength(N/m2)

Test Apparatus

Marsh cone (1 Litre)

Mud balance

Baroid filter press

Baroid filter press

pH Paper

Decantation test

Shearometer(10 min Gel Strength)

'Fresh' Mud

> 32

1.015 to 1.03

< 25

<2

8 to 10

-

L4to 10

During Excavation

-

-

< 50

< 6

8 to 11

-

-

Before PlacingConcrete

< 40 to 45

< 1.15 to 1.2

< 35 to 40

< 3

8 to 11

Generally < 3

-

toK>oo

229

Table 25 - Causes and Mitigation of Possible Defects in Replacement Piles(Sheet 1 of 4)

Defect

Hollow on the surface ofpile shaft with associatedsmall bulbous projectionsome short distance beneathhollow

Discontinuity in pile shaftwith associated largebulbous projection someshort distance beneathcavity

Soil or debris embedded inconcrete near top of pile

Possible Cause of Defect

(a) Overbreak in unstablestrata

(b) Use of doubletemporary casings andextraction of outercasing before innercasing resulting in localcavitation

(c) Intrusion of very softpeat or organic layers

Overbreak in unstable strata

(a) Overbreak in coarsegravel or fill nearground surfaceproducing sudden lossof concrete whencasing is extracted

(b) Topping up'operations, i.e.additional concretedischarged on top ofprevious lift aftercasing is removed, orinsufficientdisplacement of poorquality concrete abovethe cut-off level bytremie method.

Precautionary Measures

(a) Advancing temporarycasing ahead of bore

(b) Drilling using bentoniteslurry

(c) Use of permanentcasing

(a) Extraction of innercasing before outercasing

(a) Provision of permanentcasing

(a) Advancing temporarycasing ahead of bore

(b) Drilling using bentoniteslurry

(c) Use of permanentcasing

(a) Advancing temporarycasing ahead of bore

(b) Drilling using bentoniteslurry

(c) Use of permanentcasing

(a) Topping up' afterremoval of casingshould not be allowedand sufficient concretemust be placed toensure sound concreteat and below 'cut-off1

level.

230

Table 25 - Causes and Mitigation of Possible Defects in Replacement Piles(Sheet 2 of 4)

Defect

Soil or rock debris at baseof piles

Debris embedded in pileshaft

Local reduction in diameterof shaft of bored piles(necking) with associatedbulbs at greater depths

Local reduction in diameterof shaft of bored piles(necking) without associatedbulbs at greater depths

Discontinuities in pile shaft

Possible Cause of Defect

Dislodgement of smallblocks of soil or rockmaterial from sides of bore,sometimes caused by delayin concreting the shaft

Poor workmanship or lackof short length of temporarycasing at top of pile bore

Insufficient confinement ofconcrete in cohesive soilswith very low shearstrength

Insufficient head ofconcrete within steel casingduring extraction

(a) Low-workabilityconcrete,

(b) Premature setting ofconcrete or excessiveperiod of time betweenmixing concrete andextraction of casing

(c) Low-workabilityconcrete in lowerportion of pile shaft asa result of lack ofcontinuity in placementof concrete

(d) Aggregate interlockand raising of concretewithin casing duringextraction from use ofpoker vibrator

Precautionary Measures

(a) Concrete shaft withminimum delay

(b) Use of temporarycasing

(c) Drilling usingbentonite slurry

(a) Provision of shortlength of temporarycasing which projectssufficiently aboveground surface

(a) Problem maysometimes bealleviated by carefulslow extraction of thetemporary casing

(b) Provision of permanentcasing

(a) Adequate head andworkability of concretewithin casing

(a) Use of highworkability concretemixes

(b) .Care should be takenin hot weather

(c) Proper planning ofsupply of ready-mixconcrete; use ofretarders

(a) Proper design ofconcrete mix to ensureself-compaction,

(b) Prohibit use of pokervibrator

231

Table 25 - Causes and Mitigation of Possible Defects in Replacement Piles(Sheet 3 of 4)

Defect Possible Cause of Defect Precautionary Measures

Distortion of pile shaft Lateral movements of steelcasing during extraction

(a) Adequate groundrestraint to minimiseplant movement.

(b) Provision of adequategranular workingplatform

Containment of concretewithin cage with resultantlack of cover to rein-forcement or lack ofconcrete in bell-out.

(a) Excessive quantity ofreinforcement in cage

(a) Use of a few heavysteel sections ratherthan a large number ofclosely-spacedreinforcing bars

(b) Low-workabilityconcrete

(a) Use of high-workability concretemixes

Collapse of reinforcementcage

Inadequate design orconstruction of cage

(a) Proper design of cagewhich should besufficiently rigid andcapable of withstandingnormal site handling

Dilution of cement pasteand formation of softcement paste

Penetration of groundwaterinto body of pile because ofincorrect mix design

(a) Proper design ofconcrete mix

Excessive bleeding of waterfrom the exposed surface attop of pile

Concrete mix with a highwater-cement ratio

(a) Proper design ofconcrete mix

Weak and partiallysegregated concrete nearpile base

Significant accumulation ofgroundwater at base of boreprior to placing of firstbatch of concrete

(a) Use of tremie forconcreting

Inclusions of clay lumpswithin pile shaft

Clay lumps adhering totemporary casing which aresubsequenty displaced bythe viscous concrete andincorporated in the body ofthe pile

(a) Use of clean casing

Occasional segregation ofconcrete in pile shaft

Concrete impinging onreinforcement cage duringplacing

(a) Use of short length oftrunking to directconcrete;(NOTE : fall lengthtremie pipe must beused with raking piles)

232

Table 25 - Causes and Mitigation of Possible Defects in Replacement Piles(Sheet 4 of 4)

Defect

Segregation of concretewith dilution of cementpaste and formation of softcement paste; sometimeslayers of sand and gravelare found within body ofpile

Disintegration of concrete

Possible Cause of Defect

(a) Uncontrolled activationof trip mechanism inconcrete placers usedto place concrete inwater-filled bores

(b) Raising of tremie pipeabove surface ofconcrete eitheraccidentally or in anattempt to re-startplacing afterinterruption of freeflow of concrete downtremie

(c) Significant groundwaterflow through relativelypermeable strata

Chemical attack

Precautionary Measures

(a) Use of tremie

(a) Proper use of tremie(NOTE : tremie pipemust be water-tight anda buoyant plug ofmaterial should be usedas a separation layerbetween the first batchof concrete and wateror bentonite slurry inthe tremie)

(a) Use of permanentcasing

(a) Proper siteinvestigation includingchemical testing

Note : Table based on Thorburn & Thorburn (1977).

233

Table 26 - Interpretation of Vibration Tests on Piles

DynamicStiffness,

Kd

As expected

Low

Very low

High

Very high

High

Low

High

Low

ResonatingPile Length,

Vc/(2.Af)

As built

As built

Short

Near as built

Short

Multiplelength

Multiplelength

As built

As built

CharacteristicMobility, M0

As expected

High

Very high

Low

Low

Variable/low

Variable/high

As expected

As expected

Pile Integrity Assessment

Regular pile

Possible reduction in pile section orlower grade concrete in pile

Possible defect at depth

General oversized pile section

Possible bulb at depth

Irregular pile section in pile shaft(enlargements)

Irregular pile section in pile shaft(constrictions), or changeablequality of concrete

Regular pile with strong anchorage.Low settlement expected.

Regular pile with weak anchorage.High settlement expected.

Note : Table based on Robertson (1982).

Table 27 - Classification of Pile Damage by Dynamic Load Test Results

/3Z factor

1.0

0.8- 1.0

0.6-0.8

Below 0.6

Severity of Damage

Undamaged

Slightly damaged

Damaged

Broken

Notes: (1) Table based on Rausche & Goble (1979).(2) j8z factor is the ratio of impedance of the

given level.(3) Impedence, Z, is defined as follows :

Z = EpAp/c = Fp/v,where Ep is Young's modulus of pile,

Ap is cross-sectional area of pile,c is velocity of longitudinal stressFp is force at a given pile section,v is particle velocity

pile section above and that below a

wave through the pile,

Table 28 - Loading Schedules and Acceptance Criteria for Pile Load Tests in Hong Kong (Sheet 1 of 2)

AuthorityMax Load

inCycle 1

Max Loadin

Cycle 2

Max Loadin

Cycle 3Acceptance Criteria Remarks

GeneralSpecificationfor CivilEngineeringWorks (HongKongGovernment,1992a)

25% Qm 50% Qn 100% (1) 8 Q < 2 x 5900/oQ, and

(2) 8 > 20 mm for buildings atworking load, and 6 } 10 mmfor other structures (e.g. bridges,etc.) at working load

(1) Load increments/decrements to be in 25% of the design workingload; pile to be unloaded at the end of each cycle.

(2) Preliminary piles to be tested to not less than twice designworking load (i.e. Qmax * 2WL); working piles to be tested tonot less than 1.8 times design working load (i.e. Qmax * L8WJ.

(3) Load increments/decrements not to be applied until rate ofsettlement/rebound of pile is less than 0.1 mm in 20 minutes.

(4) Full load at each cycle to be maintained for at least 24 hoursafter rate of settlement has reduced to less than 0.1 mm perhour.

BuildingsOrdinanceOffice (BOO,1990)

100% WL 200% WL / i \ c Qmax^p D,(1) 8raax * ~£E +126 ( |nmm>

,(2) 6res *• — + 4 (in mm)

(1) Load increments/decrements to be in 50% of the design workingload; pile to be unloaded at the end of each cycle.

(2) Piles to be tested to twice design working load.

(3) Increments of load not to be applied until rate ofsettlement/rebound of pile is less than 0.05 mm in 10 minutes.

(4) Full load at cycle 2 to be maintained for at least 72 hours.

Table 28 - Loading Schedules and Acceptance Criteria for Pile Load Tests in Hong Kong (Sheet 2 of 2)

AuthorityMax Load

inCycle 1

Max Loadin

Cycle 2

Max Loadin

Cycle 3Acceptance Criteria Remarks

HousingDepartment

50%WL +100%NSF

100 %WL +200%NSF

200 %WL +300%NSF (1)

APEP-f (in mm)

(2) 6res * — or 10 mm,

whichever is smaller, if pileis embedded in soil

(3) 5res > —- or 5 mm,1UU

whichever is smaller, if pileis bearing on rock

(1) Load increments/decrements to be in 25% of the design workingload; pile to be unloaded at the end of each cycle.

(2) Pile to be tested to twice design working load; test load toinclude allowance for negative skin friction if this is expected inthe long term.

(3) Increments of load not to be applied unless rate ofsettlement/rebound of pile is less than 0.05 mm in 10 minutes.

(4) Full load at cycle 3 to be maintained for at least 24 hours afterthe rate of settlement has reduced to less than 0.1 mm per hour.

Legend : <5q Pile head settlement at failure or maximum test load£>90%Q Pile head settlement at 90% of failure or maximum test load6max Maximum pile head settlementd Pile head settlement<5res Residual (or permanent) pile head settlement upon unloading from maximum test loadQmax Maximum test loadWL Design working loadNSF Negative skin frictionLp Pile lengthAp Cross-sectional area of pileEp Young's modulus of pileD! Least lateral dimension of pile sectionD2 Diameter of pile, or diagonal width of pile where pile section is not circular

to

Notes : (1) Term "QmaxLp/ ApEp" represents the elastic compression of an end-bearing pile. Where the pile is long and resists most of the applied load in skinfriction, the elastic compression will be less than that given by the above expression.Permanent settlement of a long pile can be 'locked in1 as a result of reversal of skin friction upon removal of the test load. Limiting residualsettlement criteria are therefore inappropriate, particularly for friction piles.It is suggested that the loading schedule and acceptance criteria put forward by the General Specification for Civil Engineering Works (Hong KongGovernment, 1992a) should be adopted where possible.Suggested allowable pile head settlement at working load is for guidance only and may need to be revised for individual projects, depending on thesensitivity of the structure, elastic compression of the pile, etc.

(5) The acceptance criteria for Housing Department are based on its internal document and Yiu & Lam (1990).

(2)

(3)

(4)

236

Table 29 - Suggested Values of CASE Damping Coefficient for Different Soils

Soil Type

Sand

Silty Sand or Sandy Silt

Silt

Silty Clay and Clayey Silt

Clay

Suggested Range for jc

0.05 - 0.20

0.15-0.30

0.20 - 0.45

0.40 - 0.70

0.60- 1.10

Correlation Value of jc

0.05

0.15

0.30

0.55

1.10

Legend :jc CASE Damping Coefficient

Notes : (1) Table taken from Rausche et al (1985).(2) Correlation value of jc is the single 'best' value of jc selected for each soil

type. Using these values of jc, Rausche et al (1985) found excellentagreement between CASE Method and statically measured pile capacity in97 sites.

(3) The above recommended jc values are predominantly for end-bearing piles.

237

FIGURES

238

239

LIST OF FIGURES

Figure page

No. No.

1 Simplified Geological Map of Hong Kong 243

2 Representation of a Corestone-bearing Rock Mass 244

3 Suggested Procedures for the Choice of Pile Type for a 245Site

4 Idealised Pile Model for Wave Equation Analysis 246

5 Relationship between Nq and & 247

6 Relationship between /3 and <£' for Bored Piles in Granular 248Soils

7 Design Line for a Values for Piles Driven into Clays 249

8 Relationship between Ultimate Base Resistance and SPT N 250Value

9 Correlation between Allowable Bearing Pressure and RQD 251for a Jointed Rock Mass

10 Bearing Pressure Coefficient Ksp 252

11 Failure Modes in a Jointed Rock Mass 253

12 Load Distribution in a Rock Socket 254

13 Estimating Ultimate Shaft Resistance of a Rock Socket 255

14 Failure Mechanisms for Belled Piles in Granular Soils 256Subject to Uplift Loading

15 Failure Modes of Vertical Piles under Lateral Loading 257

16 Coefficients Kq and KC for Short Piles Subject to Lateral 258Loading

17 Ultimate Lateral Resistance of Short Piles in Granular Soils 259

18 Ultimate Lateral Resistance of Long Piles in Granular Soils 260

19 Influence Coefficients for Piles with Applied Lateral Load 261and Moment (Flexible Cap or Hinged End Condition)

240

Figure Pa§e

No. N°-

20 Influence Coefficients for Piles with Applied Lateral Load 262(Fixed against Rotation at Ground Surface)

21 Reduction Factors for Ultimate Bearing Capacity of Vertical 263Piles under Eccentric and Inclined Loads

22 Estimation of Negative Skin Friction by Effective Stress 264Method

23 Definition of Marble Quality Designation (MQD) 265

24 Bending of Piles Carrying Vertical and Horizontal Loads 266

25 Buckling of Piles 267

26 Load Transfer Analysis for a Single Pile 268

27 Closed-form Elastic Continuum Solution for the Settlement 269of a Compressible Pile

28 Depth Correction Factor for Settlement of a Deep 270Foundation

29 Horizontal Deflection of a Laterally-loaded Pile in Granular 271Soils Using the Subgrade Reaction Method

30 Analysis of Behaviour of a Laterally-loaded Pile 272Using the Elastic Continuum Method

31 Results of Model Tests on Groups of Instrumented 273Driven Piles in Granular Soils

32 Failure Mechanisms of Pile Groups 274

33 Results of Model Tests on Pile Groups in Clay 275under Compression

34 Results of Model Tests on Pile Groups and Footing 276in Granular Soil under Tension

35 Polar Efficiency Diagrams for Pile Groups under 277Eccentric and Inclined Loading

36 Determination of Distribution of Load in an 278Eccentrically-loaded Pile Group Using the 'Rivet Group'Approach

241

Figure PageNo. No.

37 Equivalent Raft Method 279

38 Group Interaction Factor for the Deflection of 280Pile Shaft and Pile Base under Axial Loading

39 Typical Variation of Group Settlement Ratio and Group 281Lateral Deflection Ratio with Number of Piles

40 Calculation of Stiffness Efficiency Factor for a Pile Group 282Loaded Vertically

41 Simplified Calculation Procedure for a Piled Foundation 283Allowing for Soil-Structure Interaction

42 Interaction of Laterally Loaded Piles Based on Elastic 284Continuum Method

43 Analysis of a Piled Raft Using the Elastic Continuum Method 285

44 Pile Head Protection Arrangement for Driven Concrete Piles 286

45 Measurement of Pile Set 287

46 Relationships between Peak Particle Velocity and Scaled 288Driving Energy

47 Possible Defects in Bored Piles due to Water-filled Voids 289in Soil

48 Detection of Pile Defects by Sonic Coring 290

49 Typical Results of a Vibration Test 291

50 Examples of Sonic Echo Test Results 292

51 Typical Arrangement of a Compression Test Using 293Kentledge

52 Typical Arrangement of a Compression Test Using 294Tension Piles

53 Typical Arrangements of a Lateral Load Test 295

54 Typical Arrangement of an Uplift Test 296

55 Typical Instrumentation Scheme for a Compression Test 297

242

Figure Page

NO. No;

56 Typical Load-Settlement Curves of Piles 298

57 Comparison of Failure Loads in Piles Estimated by Different 299Methods

58 Definition of Failure Load by Brinch Hansen's 90% Criterion 300

59 Analysis of Lateral Load Test 301

114°00'£

GUANGDONG SHENG(SHENZHEN SPECIAL ECONOMIC ZONE)

HAUHOIWAN(KEPBM)

QUATERNARY DEPOSITSLANTAU ISLANDVv

CRETACEOUS - TERTIARYSEDIMENTARY ROCKS

HONG KONG ISLANDJURASSIC / CRETACEOUSGRANITOIDS

JURASSIC / CRETACEOUSVOLCANIC/ SEDIMENTARY ROCKS

DEVONIAN - PERMIANSEDIMENTARY ROCKS

Figure 1 - Simplified Geological Map of Hong Kong

244

ft ft * ft• • II • •

Notes: (1) Figure base on Malone (1990).(2) Refer to Geoguide 3 (GCO, 1988) for classification of rock decomposition grades I to

VI.

Figure 2 - Representation of a Corestone-bearing Rock Mass

Assess Foundation Loads

Assess Ground Conditions and Types of Structures

AREPILES

NECESSARY ?

NO

Technical Considerations for Different Pile Types

GroundConditions

(Section 4.2.2)

LoadingConditions

(Section 4.2.3)

EnvironmentalConsiderations(Section 4.2.4)

^

Site and PlantConstraints

(Section 4.2.5)

Safety

(Section 4.2.6)

'

List all technically feasible pile types and rank them in order of suitability based on technical consideration

Choose Shallow

Foundation Type

Assess cost of each suitable pile type and rank them based on cost consideration

Assess construction programme for each suitable pile type and rank them based on programmeconsideration

Make overall ranking of each pile type based on technical, cost and programme considerations

SUBMIT INDIVIDUAL AND OVERALL RANKINGS OF EACH PILE TYPE TO CLIENT AND* MAKE RECOMMENDATIONS ON MOST SUITABLE PILE TYPE

Figure 3 - Suggested Procedures for the Choice of Pile Type for a Site

246

Oashpot

Spring

RdO)

(a) Actual Pile

BaseResistance

3 6K(m)

Ql J(m)

M3 o* Ioo £!

Compression at time tC(m,t)

Internal Spring

Velocity

Dashpot

G'(m)Displacement

External Spring

(b) Idealised Pile Model (c) Soil ModelBasic wave equations generally adopted for pile driving analysis are as follows :

D(m,t) = D(m,t-At) H- At V(m,t-At)C(m,t) = •D(m,t)-D(m+l,t)F(m,t) = C(m,t) K(m)V(m,t) = V(m,t-At) + [F(m-l,t) + W(m) - F(m,t) - R(m,t)] gAt/W(m)R(m,t) = [D(m,t) - D'(m,t)] K f(m) [iH-J(m) V(m,t-At)]D(m,t) = Gf(m), R(m,t) = [D(m,t) - Df(m,t)] K'(m) + J(m) Rsu(m) V(m,t-At)

With no dampingWith damping :

Legend :

m Element numbert Time

At Time intervalC(m,t) Compression of internal spring m at time tD(m,t) Displacement of element m at time tD'(m,t) Plastic displacement of external spring (i.e. the surrounding ground) m at time tF(m,t) Force in internal spring m at time t

g Acceleration caused by gravityJ(m) Soil-damping constant at element mK(m) Spring constant for internal spring mK'(m) Spring constant for external spring mR(m,t) Force exerted by external spring m on element m at time tV(m,t) Velocity of element m at time tW(m) Weight of element mG'(m) Quake for external spring m (or maximum elastic soil deformation)

.R,u(m) Ultimate static resistance of external soil spring mRd(m) Dynamic resistance of element m

Note Figure based on Poulos & Davis (1980).

Figure-4-- Idealised Pile Model for Wave Equation Analysis

247

1000

, 100

bJD

PQ

I/I

:z_

25 30 35

(degrees)

40 45

Legend :

Angle of shearing resistance of founding material

Note : Figure based on Berezantzev et al (1961).

Figure 5 - Relationship between Nq and

248

0.6

O C.5

^ 0.4-a.20e3 0.3.3•Is

1 0.2CO

0.1

0.0

—

__

—

__

• !

//

\

/

//

,

i

/

//

I/

//

/

f

\ \

—

—

—

•—

30 32 34 36 38 40

Angle of Shearing Resistance , $ (degrees)

Note ; Figure adopted from Poulo's & Davis (1980) based on interpretation ofresults given by Meyerhof (1976)>

Figure 6 - Relationship between /3 and $ for Bored Piles in Granular Soils

0.7

I05

0.30.2 0.5 1 2 3

Ratio of Undrained Shear Strength to Vertical Effective Stress, cu/av'

K)

Note : Figure based on Nowacki et al (1992).

Figure 7 - Design Line for a Values for Piles Driven into Clays

250

aOH

I• T-*Iffla

0.6

0.4

'£

P-i n o00 0.2

Coarse sand

Fine sand

Normally consolidated silt

Coarse sand

Fine sand

5 10 15

Depth in Bearing Stratum

Base Diameter

Driven Piles

Bored Piles

20

1.0

0.75

0.5

0.25

JL0.5 1.0 1.5

Base Diameter (m)

2.0 2.5

Notes : (1) Figure based on Meyerhof (1986).(2) The ultimate base resistance estimated should be multiplied by the

reduction factor,rb,when the pile base diameter exceeds 0.5 m.(3) Figure is intended for Nt value of up to about 80. Caution should be

exercised in using this Figure for higher Nj values.

Figure 8 - Relationship between Ultimate Base Resistance and SPT N Value

251

30

cj

§

PQJJio

20

10

If qa > <JC (uniaxial compressive strength ofrock), use ac instead of qa.

If RQD is fairly uniform, use average RQDwithin db = Dbwhere db= depth belowbase of foundation and Db= width offoundation.

If RQD within db= Dfe /4 is lower,use the lower RQD.

20 40 60

RQD (%)'-

80 100

Note : Figure based on Peck et al (1974).

Figure 9 - Correlation between Allowable Bearing Pressure and RQD for a JointedRock Mass

252

0.2 -

0.1 -

0 0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6 1.8 2.0

3+cd/Db

+3005d/cd

where = spacing of discontinuities= aperture of discontinuities= diameter of pile base

VALID FOR 0.05 < cd/Db < 2.0 and dd/cd < 0.02where cd t 300 mm, Db > 300 mm &

<5d < 5 mm (or <5d < 25 mm if filled with debris)

Notes : (1) The coefficient Ksp takes into account size effects and presence of discontinuities andcontains a factor of safety of at least ten against general shear failure.

(2) Allowable bearing pressure may be estimated from the strength of rock cores asfollows (Canadian Geotechnical Society, 1992);

qa = <rc Ksp d

where qa = allowable bearing pressureaverage unconfined compressive strength of rock coreempirical factor as given abovedepth factor = 1 + 0.4 Ls/Db < 3.4depth (length) of socket in rock

Db = base diameter of roek socket(3) Figure based on Canadian Geotechnical Society (1992).

qa =crc =

Ksp =d =

Ls =

Figure iu - Hearing Pressure Coefficient

253

Weak

(a) Thick Rigid LayerFlexure

Weak

(b) Thin Rigid LayerPunching

(c) Open Joints, cd < Db (d)Uniaxial Compression

Closed Joints, cd < Db

Compression Zone

Notes : (1) The ultimate end bearing capacity (qb) of foundations on a jointed rock may becalculated as follows :(a) For a thick rigid rock layer overlying a weaker rock (Figure (a)), the flexural

strength of the rock slab can be taken as equal to twice the tensile strength of theupper rock material.

(b) For a thin rigid rock layer overlying a weaker one (Figure (b)), the ultimate endbearing capacity is equal to the tensile strength of the upper rock material.

(c) For open joints and cd < Db (Figure (c)), qb - sum of unconfmed compressivestrength of affected rock columns.

(d) For closed joints, the ultimate end bearing capacity is given by the Bell Solution :

qb = c'Nc + Db7r'NT/2

where Db = foundation base width

7r'drNq

dr = foundation depth below rock surface7r' = effective unit weight of rock massN =2N I / 2

N/=tan2(45°+</>Y2)(2) For case l(d), c' and $' are the shear strength parameters for the rock mass. These

should be evaluated from insitu tests or estimated on the basis of semi-empirical failurecriterion such as the modified Hoek-Brown criterion (Hoek et al, 1992). Thefollowing correction factors should be applied to Nc and N7 for different foundationshapes:

Foundation Shape

Square

RectangularLb/Db = 2Lb/Db = 5

Circular

Correction factorforNc

1.25

1.121.05

1.20

Correction factorforNT

0.85

0.900.95

0.70

where Lb = foundation base length(3) The load acting on a pile in rock should be apportioned between the base and shaft as

shown in Figure 12. The ultimate skin friction may be estimated from Figure 13 forpreliminary design purposes.

(4) Figure based on Sowers (1979). ^_________________

Figure 11 •- Failure Modes in a Jointed Rock Mass

254

1 i 1Ep,Vp

i 1 1

soil

I rock

(Er,Vr)

i f I70

60 -

50 -

40 ~

30 -

20 -

10 -

Legend :

r0 Pile radiusVP Poisson's ratio of pilevI Poisson's ratio of rockFt Load at top of rock socket

dr

Young1 s modulus of pileYoung' s modulus of rockDepth below rock surfaceLoad at bottom of rock socket

Note : Figure based on Pells & Turner (1979).

Figure 12 - Load Distribution in a Rock Socket

3000

2000

1000

gf

fiC/l

400

300 |-

200

100

r~—r i—r~r n r

I I

~1 1 1 TT

3 4 5 10 20 30 40 50 100

Uniaxial Compressive Strength of Rock or Concrete (MPa)

to

Legend :A Lam et al (1991) Shiu & Chung (1994)

Notes ; (1) Figure based on Horvath et al (1983).(2) Correlations developed by Horvath et al (1983) are based on load tests on piles in rock with uniaxial compressive

strength of up to 40 MPa.

Figure 13 - Estimating Ultimate Shaft Resistance of a Rock Socket

256

The ultimate uplift load for a belled pile is given by :Quu = Nu Ab Ts' L

(a) For Pile in Loose Sand(Majer, 1955)

N = 2Ks(±l)tan4>'

(b) For Pile in Dense Sand(Vermeer & Sutjiadi, 1985)

Nu = 1 + 2(±l)tanc£'cos4>cv'Be

Legend :

Quit Ultimate uplift loadNu Breakout factor<£' Angle of friction of the soil

<£cv! Critical state friction angle of soil

\l/ Angle of dilation of the soilDs Diameter of shaftDb Diameter of bellBe Equivalent width of bellAt, Area of base of pileL Embedment length of pileKs Coefficient of earth pressureYs

f Effective unit weight of soil

Note : Figure based on Dickin & Leung (1990).

Figure 14 - Failure Mechanisms for Belled Piles in Granular Soils Subject to UpliftLoading

257

\ LoadH

^^PSW^

11

1

1***

Ii

/7/ *i

iii^- Centre of^ rotation

H

Free head Fixed head

(a) Short Vertical Pile under Horizontal Load

i He'

Fracture Fracture

Free head Fixed head

(b) Long Vertical Pile under Horizontal Load

Notes : (1) For soil modulus constant with depth (this applies to stiff over-consolidated clay)Stiffness factor R = VEpIp/khD (in units of length)where Eplp is the bending stiffness of the pile, D is" the width of the pile, kh iscoefficient of horizontal subgrade reaction (see Section 5.13.3(3)).

(2) For soil modulus increasing linearly with depth (this applies to normally-consolidatedclay and granular soil) ^^__Stiffness factor T = VOI/% (in units of length)where nh is the constant of horizontal subgrade reaction as given in Table 13.

(3) The criteria for behaviour as a short rigid pile or as a long flexible pile are as follows:

Pile Type

Rigid (short) piles

Flexible (long) piles

Soil Modulus

Linearly-increasing

L < 2T

L > 4T

Constant

L < 2R

L > 3.5R

(4) Figure based on Tomlinson (1994).

Figure 15 - Failure Modes of Vertical Piles under Lateral Loading

258

Fixed head

Point of application of equivalentfree-headed load

Bendingmoment

Max.

o=j[ D fm plan) (b) Shear Force (c) Bending Moment(a) Soil Reactions Diagram Diagram

10 is

z/D(d) Coefficients Kq and

roz/D

Notes : (1) The above passive pressure coefficients Kq and Kc are obtained based on the methodproposed by Brinch Hansen (1961). Unit passive resistance per unit width, pz, at depth z

p. = V*,, + C'KCZwhere av' is the effective overburden pressure at depth z, c' is the apparent cohesion ofsoil at depth z.

(2) The point of rotation (Point X) is the point at which the sum of the moment (EM) of thepassive resistance about the point of application of the horizontal load is zero,

EM = i>2-<e+z)D -:5>,-(e+z)D

z-o n z-x nThis point can be determined by a trial and adjustment process.

(3) The ultimate lateral resistance of the pile to the horizontal force Hu can be obtained bytaking moments about the point of rotation, i.e.

Hu(e+x) =£p zr-0 n z«x "

(4) An applied moment M can be replaced by a horizontal force H at a distance e above theground surface where M = H e.

(5) When the head of a pile is fixed against rotation, the equivalent height Cj above point ofvirtual fixity of a force H acting on a pile with a free head is given by

.zf is the depth from the ground surface to point of virtual fixity. American ConcreteInstitute (1980) recommended that zf should be taken as 1.4R for stiff, over-consolidatedclays and 1.8T for normally consolidated clays, granular soils and silts, and peat.

(6) Figure based on Tomlinson (1994).

Figure 16 - Coefficients Kq and KC for Short Piles Subject to Lateral Loading

259

HU

e ILII

i

Deflection

PL

3D7s'LKp M*

Soi/react/on

(a) Free Head

Bendingmoment

m

Deflection

16 20

3DTs'LKp

BendingSoil reaction moment

(b) Fixed Head

Notes : (1) For free-headed short piles in granular soils (see definition in Figure 15),

Hu = 0.5 D L3 Kp <ys'/(e + L)

where Kp = Rankine's coefficient of passive pressure= (1 + sin<£')/(l -sin^1)

D = width of the pile$ = angle of shearing resistance of soilY ' = effective unit weight of soil

(2) For fixed-headed short piles in granular soils (see definition in Figure 15),

The above equation is valid only when the maximum bending moment M^x whichdevelops at the pile head is less than the ultimate moment of resistance Mu of the pileat this point. The bending moment is given by

(3) The ultimate lateral resistance for free-headed and fixed-headed short piles in granularsoils can also be determined by the graph above.

( 4 ) Figure based on Tomlinson (1994). . . , • . . , . , ' . . • . . ; • , . . , ' . ' . ' , . . . . . . • . .- ,

Figure 17 - Ultimate Lateral Resistance of Short Piles in Granular Soils

260

BendingDeflection Soil Reaction Moment

(a) Free Head

Deflection Soil Reaction

(b) Fixed Head

1000 [-Mie

100

Fixed head

Free head

1.0 10 100 1000 10000

M U / D 4 Y S K P

Notes : (1) For free-headed long piles in granular soils (see definition in Figure 15),Mmax = H(e + 0.67f*)

where f* = 0.82VH/(7s'DKp)Mmax = maximum moment of the pile shaft and is equal to the ultimate

moment resistance of the pile shaft (Mu) at ultimate lateral load, Hu

D = width of pile in the direction of rotationKp = Rankine's coefficient of passive pressure

= (1 + sin <£')/(! -sin0')$' = angle of shearing resistance of soil7S' = effective unit weight of soil

(2) For fixed-headed long piles in granular soils, the maximum bending moment occurs atthe pile head and at the ultimate load. It is equal to the ultimate moment of resistanceof pile shaft.

M,max = H(e + 0.67f*)/2For a pile of uniform cross-section, the ultimate value of lateral load Hu is given bytaking the Mmax as the ultimate moment of resistance of the pile Mu.

(3) The ultimate lateral resistance for frpe-headed and fixed-headed long piles in granularsoils can also be determined from the graph above.

(4) Figure based on Tomlinson (1994).

Figure 18 - Ultimate Lateral Resistance of Long Piles in Granular Soils

0

1

z 2

T

3

4

0

1

z 2

T

3

4

Notes : (1) T =(Tat

(2) Obtaresp<

(3) Figu

-1 0 1 2 ^ - 3Deflfora

-LT

Uetlfora

JL _

ection coeffpplicd mom

j

= 2 ""

ectionppliec

/

coettlaterj

"IA

icient (ent(N<

^4,5^cient(il force

k5*M

^ J

L^

:10

*fc><BL

o

^

r

2

L

H

z

L !

M y-6M

J

I/

F6 (|~&*

-1 0 1 2 ,3

Deflection Coefficient, F5

le 13). P

tin coefficientactively usingre based on I*

-0.1-p ^^-*-"{) 0.2 0.4 0.6 0.8 <— 1.0Moment coefficient (F^for applied moment (M)

T

io|\

Morfora

L _T

4,

=2 r

/ >

5

nent«ippliec

'5

~^~ — ~

>efficilatera

-..:_•_

•^- ,

Bnt(FN

1 force

j~

^*

^-^

J®

. ^.

._.___

:

^

----- -

-

^

/

L

H

z

L ,

M r-

I;FM(

)"7

=FM(Fi

V

M)

-MH

7777

rr>

/

z

Shefor E

•^•^ -0.6 -0.4 -0.2 0Shefor

-, MIcrj77Z7

ircoeippliec

arcoepElieJ

X

L

ficienti latera

•1

Ticien.mom

/

- L . .V

(Fv)J load

\

t(Fv)ent^

?^

(H)

lo^5

L,.

, . . •

^^=

^

^

-1i

VH

LT

--^

4

t|

|

= FV(I

= 2

L

0

-^^0 0.2 0.4 0.6 0.8 -0.8 -0.4 0 0.4 0.8

"°gl Moment Coefficient, FM Shear Coefficient, Fv

where Eplp = bending stiffness of pile and i

s F§, FM and Fv at appropriate depths desirethe given formulae.

JAVFAC (1982).

lh - C

d and

onstant of horizontal subgrade reaction

compute deflection, moment and shear

to

Figure 19 - Influence Coefficients for Piles with Applied Lateral Load and Moment (Flexible Cap or Hinged End Condition)

262

- H _ rfrt

zT

_._; Deflection coefficient (Fs) _ L

i for applied lateral load (H) j

-0.2 0.2 0.4 0.6 0.8

Deflection Coefficient, Fg

1.0

zT

Moment coefficient (FM)for applied lateral force (H)

/

MH=FM(HT> ~ " *

-1.0 -0.8 -0.6 -0.4 -0.2 0

Moment Coefficient, FM

0.2

Notes : (1) T = 54'(^p/niJ where Eplp = bending stiffness of pile and % =constant of horizontal subgrade reaction (Table 13).

(2) Obtain coefficients F$ and FM at appropriate depths and computedeflection and moment respectively using the given formulae.

(3) Maximum shear occurs at top of pile and is equal to the applied load H.(4) Figure based on NAVF AC (1982).

Figure 20 - Influence Coefficients for Piles with Applied Lateral Load (Fixed againstRotation at Ground Surface)

263

e/D0 0.2 0.5 1 5 oo

20° 40° 60° 80° 90°

Angle tan'1 (e/D)

(a) Eccentricity Factor

0 20° 40° 60° 80*90°

Angle of Inclination from Vertical, <XL

(b) Inclination Factor

Legend:

dcD

Measured values in loose sandMeasured values in soft clayMeasured values in clay overlying sand (d</D = 0.5)Theoretical relationshipEccentricity from centre of pileAngle of inclination from verticalThickness of clay layerPile width

Notes : (1) = r er iQ0

where Qv = vertical component of the ultimate eccentric inclinedload

Q0 = ultimate concentric vertical loadre = reduction factor for eccentricityT{ = reduction factor for inclination of load from vertical

(2) The values of re and q may be obtained from Figures (a) and (b) aboveor from the following equations :

• • • tan4 (e/D) .-re = UFor granular soil,

For clay,

90°= (1 - «L/90b) 2

tan"1 (e/D)

.J

90C

= COS

(3) Figure based on Meyerhof (1986).

Figure 21 - Reduction Factors for Ultimate Bearing Capacity of Vertical Piles underEccentric and Inclined Loads

264

Actual distribution

Calculated distributionof load in pile

Notes : (1) The peak negative skin friction in granular soils and cohesive soils is determined as forpositive skin friction. Using the effective stress approach, the ultimate negative skinfriction can be estimated as follows :

f n = . / 3 e r v '

where f, = ultimate negative skin friction<7v

f = effective vertical stress]8 =. empirical factor obtained from full scale load tests or based on the

following table for preliminary design purposes

Soil Type

Clay

Silt

Sand

, $• (after Garlanger et al> 1973)

0.20-0.25

0.25 - 0.35

0.35 -0.50

(2) The neutral point may be taken to be at the pile base for an end-bearing pile that hasbeen installed through a thick layer of soft clay down to rock or to a stratum with highbearing capacity. For a pile whose tip could settle when loaded, the ratio of the depthof the neutral point to the length of the pile in compressible strata may be roughlyapproximated as 0.75 (NAVFAC, 1982).

Figure 22 - Estimation of Negative Skin Friction by Effective Stress Method

265

§

<D

P

I

100

L2 (mPD)

I a

75

Zero marble rockcore either cavity ordecomposednon-marble rock —

50

25

10

D

LI " LI

Marble Rock Recovery Ratio (MR) =

where L2 - Lj usually = 5 m

MQD = Average RQD x MR

L I - L I -

MarbleClass

II

in

rv

2 3

Total Cavity Height (m)

Note : At the rockhead, where the top section is shorter than 5 m but longer thanor equal to 3 m, the MQD is calculated for the actual length and designatedas a full 5 m section. If the top section is shorter than 3 in, it is to begrouped into the section below.

Figure 23 - Definition of Marble Quality Designation (MQD)

266

H

(a) Vertical Loading ona Out-of-plumb Pile

(b) Applied andInduced Loadingon Pile

(c) EquivalentLoadingon Pile

H =

ef =

M = He'

Legend:

e1 Effective eccentricity of loadP Applied vertical loadH Induced horizontal load due to non-vertically of pilee Eccentricity of load applicationer Free length of pile above ground levelM Moment on pile/?' Inclination of pile

Notes: (1) The analysis of a pile subject to moment and lateral load can be madeusing Figure 19 or 20 as appropriate.

(2) The depth of any near-surface weak material should be included as partof the eccentricity e f .

Figure 24 - Bending of Piles Carrying Vertical and Horizontal Loads

267

Applied Load Applied Load

o (Critical length under^ lateral loading)

(a) Actual Pile (b) Equivalent Cantilever

For piles with a free end, Pcr =

where L = —G

2/7 1/4

for soils with a constant Kh

« 4El.p P

1/5

For piles with a fixed end, Pcr = p p

for soils with a linearly increasing Kh

Legend :

er

LDGcG*G

Critical buckling loadYoung's modulus of pileMoment of inertia of pileFree length of pile above groundCritical pile length for lateral loadingTotal pile lengthPile diameterMean value of G* over Lc

Shear modulus of soilPoisson's ratio of soilModulus of horizontal subgrade reactionConstant of horizontal subgrade reaction

Note : Figure based on Fleming et al (1992).

Figure 25 - Buckling of Piles

268

F.CD

Q)©

Pile

0

©

| "

L iT, |

-^jh5l^JQ,

-®-h5^jo,

iQi© ^

" fix J' *Oi ^

ShearStress,

Ti

Qi&Q} + 1 = Load acting on element it{ = Shear stress on element i6j = Movement at the middle of

element i

Li = Length of element i© = Element number

Mean Displacement,

Typical Assumption of ShearStress-Displacement Relationshipfor Element i

Procedures :

L Compute tip load Qn+i corresponding to a given base movement d^ based on assumed base stress-displacement relationship.

2. Estimate midpoint movement 5n for bottom element n; for the first trial, assume 6n = <5b.

3. Given 6n, the shear stress rn can be determined for given shear stress-displacement relationship.

4. Calculate Qn : Qn = Qn+1 + j (rn x pn x Ln), where pn is the perimeter length

5. Assuming a linear distribution of load along the pile element, compute the elastic defonnation 5das forthe bottom half of the element :

A -_ 0>5 { (Qn + Qn+l)/2 + Qn+1elas —6. Compute new 6n : 8n = 6b 4- 5elas

7. Compare new 5n with that initially assumed in step 2. Adjust and repeat analysis until specifiedtolerance is achieved.

8. When required convergence is achieved, proceed to next element up and repeat the procedure.Continue until the load at the top of the pile, Qj, is computed corresponding to a given value of 6t>.

9. Repeat the calculation procedure using a different assumed 8^ and establish the complete load-settlement relationship at the top of the pile.

Note : Method based on Coyle & Reese (1966).

Figure 26 - Load Transfer Analysis for a Single Pile

269

Pile

(a) Friction Pile (b) End-bearing Pile

Assumed Variation in Modulus with Depth

For an applied load Pt, the pile head settlement (5t) of a compressible pile is given by thefollowing approximate closed form solution :

4 ryr 2?rp tanh(ftL) LPt _ d-"s)£ f W ~

1 4 7?r tanh(^L) Lr0GL

where7?r = rb/r0 (r0 is the radius of pile shaft & rb is radius of pile base)£ = GL/Gb (GL & Gb is shear modulus of soil at depth L & pile base respectively)P = GL/2/GL (Rate of variation of shear modulus of soil with depth)X = Ep/GL (pile stiffness ratio)

T r 2 i 1 / 2 LM = [ —Z£ ] —

f = In { [ 0.25 + (2.5p(l - ?s) - 0.25)£ ] L/r0 }vs = Poisson's ratio of soil

The settlement profile with depth may be approximated as :

5 = 5b cosh 0*(L-z)) where 5b = b ^s— , Pb = Load at pile base

For a non-circular pile, Ep may be taken as : Ep = (E A)p/7r r02

Where the slenderness ratio L/D is < 0.25 (Ep/GL)*, the pile may be treated as effectively rigidand the pile head stiffness is given by :

Pt _ _ * 5 r ' ,(6 tr0GL) (l-vs)£ r0

Where L/D is > 1.5 (Ep/GL)1/2, the pile may be treated as infinitely long. In this case, theeffective pile head stiffness is given by :

_ _(6t r0 GL)

or Pt * 2 p r0 (Ep GLJ1/2

For this case, GL is the soil shear modulus at the bottom of the active pile length Lac, whereLac = 3r0(EpGL)1/2.. ./

Note : Method based on Fleming et al (1992).

Figure 27 - Closed-form Elastic Continuum Solution for the Settlement of aCompressible Pile

Ti

N>00

&9

a.§Tl8

?(/)

s

Oi-bP

51

I

^rB-S^

B'»

21 §"<JQ aCri ' J"3

5?> P?£§ yc/3 CLs PPo ^^g OP< C«8L~

|1I

O 3

o H a -i* 8: ff I.

OTQ ?•- O

I! a

Settlement of Deep LoadSettlement of Corresponding Surface Load

pLn

pbo

^ v

Oo

II

271

F

12

S/— S

& 10

3-s $€ .to

Jl 6

e1 4

CO

o

1 2

0

iFixed-headed

f ~

1 , Free-headed

/ H^nft y ,t r

- l\/ /v( /' /\\^^-^/ 1^ ^ — ~ — "" — ~•\ ^^^^^^^

i "" ^ i - — i i — —

e/L«2.0

1.5

1.00.80.60.4

IK2 4 6 8 10

Dimensionless Length, L/T

For piles in granular soils, define characteristic length T = (Eplp/nh)1/5

The criteria governing pile behaviour are as follows :

Rigid piles Flexible piles

Free-head conditions L < 2 T L > 4 T

Fixed-head conditions L < 2 T L > 4 T

Dr a free-headed rigid18H(1 -f 1,33 e/L)

•nilp A, —pile, 5h ^^

2, HFor a fixed-headed rigid pile, 5h = -yj —

For a free-headed flexib

For a fixed-headed flexi

Legend

YP

Note :

Icnilc 5 2.4 H ( l + 0.67 e/T)1C pllC, Oh . )3/5(EnI )2/5

blCpllC, V (nh)i/5(EpIp)2/5

Bending stiffness of pile H Horizontal load on pileLateral deflection at ground surface L Embedded length of pileConstant of modulus variation e Eccentricity of load application as

measured from ground level

Method based on Broms (1964b).

Figure 29 - Horizontal Deflection of a Laterally-loaded Pile in Granular Soils Usingthe Subgrade Reaction Method

272

M

Free-headed piles

(E/GC)1/7

H

\l/7

e =

0.27H + 0.3M(Lc/2) (Lc/2)2

0.3H + 0.8(pc')1/2M

PC' Gc (V2)2 (V2)3

The maximum moment for a pile under a lateral load Hoccurs at a depth between Lc/4 (for homogeneous soil) and Lc/3(for soil with stiffness proportional to depth). The value of themaximum bending moment Mmax may be approximated usingthe following expression :

0.1 H L

»

Pile

j

—

7"/

^\\

L

i

jTo

^ /

c

f

Fixed-headed pilesIn this case, 0 equals zero and the fixing moment Mf and lateral deflection (6^) are given

by the following expression :

0.375 H (L 12)

(Pe')«\ l /2

(EP/GC)1/7

0.27 - 0.11 HM/2 (Lc/2)

The lateral deflection of a fixed-headed pile is approximately half of that of a correspondingfree-headed pile.

Legend

Sh0G,

PC'G*

"sGHM

Lateral pile deflection at ground surfacePile rotation at ground surfaceCharacteristic shear modulus, i.e. average value of G* over the critical length Lc ofthe pileCritical pile length for lateral loading = 2 r0 (Epe/Gc)

2/7

Equivalent Young's modulus of pile = (EpIp)/(7rr04/4)

Degree of homogeneity over critical length Lc = G^/CcG ( l + 3V4)Value of G* at depth of Lc/4Poisson's ratio of soilShear modulus of soilHorizontal loadBending momentBending stiffness of pilePile radius

Note Method based on Randolph (1981a).

Figure 30 - Analysis of Behaviour of a Laterally-loaded Pile Using the ElasticContinuum Method

273

a

I

3.0 r

2.5 -

2.0

1.5

1.0

0.5

//

Shaft efficiency

9 - Pile groupTotal efficiency with pile cap

Total efficiency

Point efficiency

(average of tests)

2 3 4 5 6 7

Pile Spacing / Pile Diameter

Notes: (1) Figure based on Vesic (1969).(2) Efficiency denotes the ratio of ultimate load capacity of a pile group to

the sum of ultimate load capacities of individual piles in the group.Shaft efficiency denotes the above ratio in terms of shaft resistanceonly. Point efficiency denotes the ratio in terms of base resistanceonly.

(3) Vesic (1969) noted that in view of the range of scatter of individual testresults, there was probably no meaning in the apparent trend towardslower point efficiency at large pile spacings.

Figure 31 - Results of Model Tests on Groups of Instrumented Driven Piles inGranular Soils

Piles-[SkinI friction

End bearingstress

friction

Surface ofassumedfailure bolock

End bearingstress

>ck

h •i i

'if~^| j[ ij i

A \ \

Skinfriction

End bearingstress

to

(a) Single Pile Failure (b) Failure of Rows of Piles (c) Block Failure

Note : In assessing the ultimate base capacity of a block failure in granular soils, the effective self weight (W) of the soil abovethe founding level may be allowed for (Fleming et al, 1992).

Figure 32 - Failure Mechanisms of Pile Groups

i

CD

INJo

I

00 GO*fl H OQ Group Reduction Factor

to

276

1.0

0.8

0.6

0.4

0.2

PH

§

I

Dense Sand

• L/D = 3g | 2 footings

X L/D = 20 2 piles

A L/D = 3 1

+ L/D = 20 4 piles

Ratio of Spacing to Width of Pile

Legend :

L Length of pileD Pile width . p , | Theoretical relationships

Note : Figure based on Meyerhof & Adams (1968).

Figure 34 - Results of Model Tests on Pile Groups and Footings in Granular Soilunder Tension

277

Eccentricity Ratio y = 0

Eccentricity Ratio — = 0.8

90°0.2 0.4 0.6 0.8 1.0

3C+-4

O

O

oc

0.2 0.4 0.6 0.8 1.0

Group Reduction Factor for Horizontal Loading

CLAY

SAND

77777"

//////

Legend :

Eccentricity of applied load from centroid of pile groupAngle of inclination of applied loadThickness of clay stratumLength of pile

Notes : (1) Figure based on Meyerhof & Purkayastha (1985).(2) These model test results form a consistent set of data on the relative

effect of eccentricity and inclination of the applied load. Therecommended group reduction factors given in Sections 6.3.2, 6.3.3 &6.3.5 for concentric and vertical loading (i.e. e = 0 & aL = 0) shouldbe scaled using the ratio deduced from this Figure to take account ofload eccentricity and inclination effects.

Figure 35 - Polar Efficiency Diagrams for Pile Groups under Eccentric and InclinedLoading

278

Pi = QT/np + (My*.Xi)/Iy + (Mx*.yi)/Ix

and

Mv

where

Pi

QT

nP

MX,My

Ix,Iy

MX",My* =

- My.Ixy/Iy

My - MX.IXy/IX

1 - IXy2/(Ix.Iy)

axial load on an individual pile i

total vertical load acting at the centroid of the pile group

number of piles in the group

moment about centroid of pile group with respect to the x and y axesrespectively

moment of inertia of pile group with respect to x and y axes

respectively

product of inertia of pile group about the centroid

distance of pile i from y and x axis respectively

effective moment with respect to x and y axes respectively, taking intoaccount the symmetry of the pile layout.

For a symmetrical pile group layout, Ixy = 0 and Mx*, My* = Mx, My

In this case, the above expression becomes :

' QT My* MxYiP. -•M — I*

Notes : The assumptions made in this method are :(a) pile cap is perfectly rigid,(b) pile heads are hinged to the pile cap and no bending moment is transmitted

from the pile cap to the piles, and(c) piles are vertical.

Figure 36 - Determination of Distribution of Load in an Eccentrically-loaded PileGroup Using the'Rivet Group1 Approach

Spread of load at 1 in 4

Base of equivalentraft foundation —-

(a) Group of Piles SupportedPredominantly by Skin Friction

(b) Group of Piles Driven ThroughSoft Clay to Combined SkinFriction and End Bearing inDense Granular Soil

(c) Group of Piles Supported by EndBearing on Hard Rock Stratum

Note : Figure taken from Tomlinson (1994),

Figure 37 - Equivalent Raft Method

280

Surface settlement profiles Original ground surfacefor individual piles —v ^/

Combined surfacesettlement profile

Piles

JL-Jb

Piles

If the stiffness of a single pile under a given form of loading is KV, then a vertical load Pwill give rise to deformation Sv given by

KV

If two identical piles are each subjected to a load P, then each pile will deform by anamount 5V given by

_ (1 + a ' ) PKV

where a' is the interaction factor.

For axial loading, the interaction between the pile shafts and the pile bases can be treatedseparately :

Pile Shaftln(rm/sp)

where a =

rm =

r =sp =

shaft interaction factor for a pile at a distance sp from the pile beingloadedIn(rm/r0)maximum radius of influence of pile under axial loading; empirically thisis of the order of the pile lengthpile radiuscentre-to-centre spacing of pile

Pile Base

7T.Sr

where % is base interaction factor for a pile at a distance sp from the pile being loaded

Note : Method given by Fleming et al (1992) based on solution of Randolph &Wroth (1979).

Figure 38 - Group Interaction Factor for the Deflection of Pile Shaft and Pile Baseunder Axial Loading

281

21

5i3C

00o,

1O 5

flexible pile cap(uniform load)

rigid pile cap

sp /D=*3X=1000

L/D-25

7 9 11VnTp

(a) Rgs for p = 1 (b) Rgs forp = 0.5

17

a 9a

O. '

1O

10

3 - 5 7 9

( d ) R h forp'= 0.5

Legend

np

G*

DGL

G*Lc/EP

Number of piles in the group pGa+3?s/4) PC'Pile radius GPile length Lc

Poisson's ratio of soil Gc

Pile diameter sp

Value of G at depth L GL/2

Value of G* at depth Lc/4 XYoung's modulus of pile

Variation of soil modulus with depth = GL/2/GL

Degree of homogeneity over Lc - G*Lc/4/Gc

Shear modulus of soilCritical pile length for lateral loadingAverage value of G* over Lc

Pile centre spacingValue of G at depth L/2Pile stiffness ratio (= Ep/GL)

Note : Figure based on Fleming et al (1992).

Figure 39 - Typical Variation of Group Settlement Ratio and Group LateralDeflection Ratio with Number of Piles

282

0.6~

0.59~c3

i\ CO+f 0.58-H A CT8 "-57"

a A C£0.56-

pq 0.55-^» A £A& 0.54-JTJ r» *<2

.£> 0.53"o

• *H A CO5tt 0.52U3 A <t_" U.5 r

n <U.50

.lj

r/j 11-*2 i-iO 1 AC..03 1.05-

fert 1 _

.2O A QC_<D u.yj

O A A _O u.y•tj(H 0 »^\-53 u-5:3

g A 0 _

W /\ «r;_

07-

v\

/

j/

/

//^e*r.-^^

X ,^ss^^^>

X^^

20 40 60 80Slenderness Ratio, L / D

(a) Base Value

^Ks\

/

»^Pois

\s-

son's t^/\

X

itio, Vp

<Ss

\X

^

^

stif

_^S^

^

Spaciu'N**>_ i

Messrs

^ —m»f

*nom(.

g ratio

itio, E

^— — -

^-)geneit

, sp/r

100

p/G L

***^

y^p

)

^— — .0 0.2 0.4 0.6 0.8

Poisson's Ratio, and Homogeneity Factor, pi

" i i i - • i - i ' i r — 1 12 4 6 8 10 12

Spacing Ratio, sp / Di , , , i i i 1 12 2.4 2.8 3.2 3.6 4

_a • Log10 (Stiffness Ratio, Ep / GL)Kg — lip

(b) Correction Factors

Legend :

Ep Young's Modulus of PileRg Stiffness efficiency factora Exponent for stiffness efficiency factorL Length of pileD Pile diameterVP Poisson's ratio of pilep Rate of variation of shear modulus of soil with depth (homogeneity factor)sp Pile centre-to-centre spacingGL Shear modulus of soil at pile basenp Number of piles in a group

Notes: (1) Figure based on Fleming et al"-( 1992).(2) A base value of efficiency exponent, a, should first be chosen from Figure (a).

Correction factors are then selected from Figure (b) and multiplied by the base value.

Figure 40 - Calculation of Stiffness Efficiency Factor for a Pile Group LoadedVertically

283

Initial structural analysis with springs modellingthe piles with assumed stiffnesses (e.g. thatcorresponding to single pile)

(2) Apply pile loads (which may be converted topressure) from step 0 for settlement analysis

i Calculate settlements from a foundation analysis

@ Repeat structural analysis with spring stiffnessesassessed from previous step

i Foundation settlement analysis with pile loadscalculated from step ©

2) Comparesettlement from

step (3) with those from previoussettlement analysis :

convergence ?

No

Stop

Notes : (1) Figure based on Chan & Davies (1984).(2) Allowance may be made in structural analysis for yielding of piles

where appropriate. . . . ' . . • • . . . • • . . . - . . .;

Figure 41.- Simplified Calculation Procedure for a Piled Foundation Allowing forSoil-Structure Interaction

284

If the stiffness of a single pile under a given form of loading is KL, then ahorizontal load H will give rise to deformation Sh given by :

Sh = H / KL

If two identical piles are each subjected to a load H, then each pile will deformby an amount Sh given by :

6h = (1 + a1) H / KL

For Fixed-Headed Pilesa' = 0.6 PC' (Ep/Gc)1/7 r0 (1 + cos2co) / sp

At close spacings, the above expression over-estimates the amount ofinteraction. When the calculated value of a1 exceeds 0.33, the value should bereplaced by the expression 1 - 2 / (27a')1/2

For Free-Headed Pilesa' = 0.4 PC' (Ep/Gc)1/? r0 (1 + cos2co) / sp

PileB

Definition of Departure Angle co

Legend :

af Interaction factor for deflection of pilesPc1 Degree of homogeneity (= G^/Gc)G* G(l + 3V4)G Shear modulus of soilLc Critical pile length for lateral loading = 2 r0 (Epe/Gc)

2/7

G*Lc/4 Value of G* at a depth of Lc/4Gc Average value of G* over Lc

co Angle of departure that the pile makes with the direction of loadingv$ Poisson's ratio of soilr0 Pile radius

Epe Equivalent Young's modulus of pile = (EpIp)/(7rr04/4)

Note : Method based on Randolph (1981a) & Randolph (1990).

Figure 42 - Interaction of Laterally Loaded Piles Based on Elastic Continuum Method

285

0.8

0.4

°-2

Poulos & Davis (1980)Approximate analysis byFleming et al (1992)

L/D = 25 (?s = 0)

L/D = 25 (?s = 0.5)

L/D = 10 (?s = 0.5)

1 2 A $ a 1 0

For a piled raft where the raft bears on a competent stratum, the approach of combining theseparate stiffnesses of the raft and the pile group using the elastic continuum method is based on the useof average interaction factors, acp, between the pile and the pile raft (or cap).

The overall foundation stiffness, Kf, is given by the following expression :

Kf = (1 - c/cp. Kc/Kg)°

The proportion of load carried by the pile cap (Pc) and the pile group (Pg) is given by :Kc(l - acp)

Pc+Pg

Legend

Kg

Kc

I

Stiffness of pile group_ 2G(b.l)1/2

ness o pi e cap - l ^_^

Influence factor, see Poulos & Davis (1974) or BSI (1986)_ ln(rm/rc)

In(rm/r0)Radius of influence of pile « length of pile

Average i

bD1G"sLnp

A-cap

Equivalent radius of the pile cap for each pile =

Radius of pileWidth of raftPile diameterLength of raftShear modulus of soilPoisson's ratio of soilLength of pileNumber of pilesArea of pile cap

cap

p

Note : Method based on Fleming et al (1992).

Figure 43 - Analysis of a Piled Raft Using the Elastic Continuum Method

p

o

Io"

OPCD

iBs>1-1O3.

POooCD

CD</}

Hammer Components

00ON

CD

287

Card held by clamps,or paper stuck toface of pile

(a) Arrangement for Measurement of Pile Set

Elasticrebound

— Set

Card

(b) Typical Record of Pile Set in Hong Kong

Note : Figure (a) based on Whitaker (1976).

Figure 45 - Measurement of Pile Set

288

Peak

Par

ticle

Vel

ocity

(mm

/sec

)H

^l

)-*

N

>

LO

L

/l

Ot— '

NJ

U>

LJi

O

OO

O

O

(b)// /

///y//1/

/t

I/

/

Vs>

(e)/

/

(a) Wiss ((b) Wiss ((c) Wiss ((d) Attew((e) Brenne

'//

//

/

1967) -clay1967) -wet sand1967) -dry sand^11 & Farmer (1973) - sand & gravel, silt,;r & Chittikuladilok (1975) - clayey sand

/'

clayor stiff clay

0 20 30 50 100 200 400

/ Energy (in Joules) / Distance (m)

Notes : (1) Criteria (a) to (c) relate to seismic distance, i.e. distance from pile tipto point of measurement

(2) Criteria (d) & (e) relate to the horizontal distance between the pile axisand the point of measurement.

(3) Criteria (a) to (d) relate to the vertical component of velocity whereascriterion (e) relates to the resultant velocity.

Figure 46 - Relationships between Peak Particle Velocity and Scaled Driving Energy

Slurry

(a)Slurry filled cavityformed outside the

casing

(b)Pile concreted, casinglifted in cavity under

pressure

(c)Casing is lifted higher,

concrete slumps intothe void, contaminatedslurry flows into pile

00

Note ; Figure taken from Sliwinski & Fleming (1984).

Figure 47 - Possible Defects in Bored Piles due to Water-filled Voids in Soil

290

R

(a) Horizontal Test (b) Influence of an Irregularity (c) Inclined Test

(d) Fan-shaped Test (e) Zone of Influence(f) Irregularity near

Pile Heart

Note : Figure taken from Tijou (1984),

Figure 48 - Detection of Pile Defects by Sonic Coring

291

Signal proportional to Vt

Velocity transducer

(a) Schematic Arrangement in a Vibration Test

- Frequencyof first

resonance

Frequency, f (Hz)

(b) Idealized Response Curve from a Vibration Test

Legend :

Lj-es Resonating length = Vc/2Af PmVc Wave velocity in concrete QmK<1 Dynamic stiffness of pile head M0

Dc Damping factor = Pm/Qm PCAc Cross-sectional area of concrete

Mobility at resonance (peak)Mobility at anti-resonance (trough)Characteristic mobility = l/(pcVcAc)Density of concrete

Figure 49 - Typical Results of a Vibration Test

292

& -v /•

1 IV(a) No Echo

Echo

I V<D \/

\S

(b) Echo from Free Surface

^1>

Echo

v(c) Echo from Fixed Surface

Echo

(d) Echo from Intermediate Surface

Time (ms)

Time (ms)

Time (ms)

Time (ms)

Time (ms)

(e) Overshoot and Ringing Caused by Imperfect Deconvolution

Note : Figure taken from Ellway (1987);

Figure 50 - Examples of Sonic Echo Test Results

293

Angle lacingand tie bolt J

Universal beams

(U

Universal beamClear space

Cap cast on tohead of test pile

1.3 m minimum

Note : Figure based on Tomlinson (1994).

Figure 51 - Typical Arrangement of a Compression Test Using Kentledge

294

^Universal beams —

Bracing

Load cellHydraulic jack Dial gauge supports

Ball joint

Test pile

Tensionmembers

-D

Tension (anchor) piles

Clear space

Minimum spacing2m or 3D whichever is greater

Note : Figure based on Tomlinson (1994).

Figure 52 - Typical Arrangement of a Compression Test Using Tension Piles

295

Reference beam- Dial gauge

Steel strutHydraulic jack

(a) Reaction Piles

Bearing plate

^cking\ r

Hydraulic jack

Test plate

Deadman

x —

-aft-Reference beam

Test pile

(b) Deadman

Weights

Platform

Beaming plates

Test plate

Hydraulic jack

Dial gauge

I \ s— Reference beam

Test pile

(c) Weighted Platform

Note : Load cell with appropriate plates may be inserted between test plate andhydraulic jack.

Figure 53 - Typical Arrangements of a Lateral Load Test

Steelbearing plate

Reaction pileor cribbing

Test Beam (s)

Clearancefor pile

movement

Test pile

Minimum spacing2 m or 3D whichever is greater

Reaction beam

- Steel plate

Hydraulic jack ram

Steel bearing plate

Tension connection

Dial gauge

' j£*— Reference beam

Steel BearingPlate

toVDON

Note : Figure based on Tomlinson (1994).

Figure 54 - Typical Arrangement of an Uplift Test

297

Base load cell

- Rod extensometers

v- Strain gauges

(a) Typical Instrumentation Scheme

Read cables

Plastic tube

Pressure pad

Load cell unit

Soft membrane

Steel plate

Bonding bars

Airline

Protective concrete

Skirt

Inflatable tube

Recess

Price & Wardle's (1983) type

(b) Details of Typical Base Load Cell

Legend :

I Vertical strain gauge•-< Horizontal strain gauge

Figure 55 - Typical Instrumentation Scheme for a Compression Test

of—KCD

(n>ui0\iH^<*T3

j— i.O

opO-'

00<p

CD

or.PGOCDP-

OP

Ho

00O

^

S3O CD

1CD

B.Od

CD

*3

I

3CD

So

CD(JO

O o*B. B.

a§

005±.ajn

299

3000

2500

2000

•so

1500

1000

500

Chin (1970) [2395]

Brinch Hansen (1963) [2050]

Yiu& Lam (1990) [1982]

Davisson(1972)[1918]

10

D = 0.305 m

Soft Clay

Clayey Silt

Silt

£00

S;

J* S </ <*?

Ep = 29.65 x l O 6 kN/m2

20 30

Settlement (mm)

40 50

Legend :

Young ' s modulus of pile D Pile width

Notes : (1) Numbers in [ ] are the ultimate loads estimated by the method given inthe reference.

(2) Figure based on Fellenius (1980).

Figure 57 - Comparison of Failure Loads in Piles Estimated by Different Methods

300

2500

2000 -

1500 -

1000 -

500 -

10 20 30

Settlement (mm)

40 50

Ultimate load, Qult, in accordance with the 90% criterion of Brinch Hansen(1963) is given by the following :

Qult = 2050 kN where gSettlement at 90% Q = 2ult

Figure 58 - Definition of Failure Load by Brinch Hansen1 s 90% Criterion

301

Load •

Apparent pointof rotation —*

&/6 = butt slope

(a) Definition of Apparent Point of Rotation

Load Load

''/If/- Shaft failure point

(depth of apparentpoint of rotationremains constant)

Rigid body rotationofshaft

(depth of apparentpoint of rotationremains constant)

(b) Conditions for Constant Depth of Apparent Point of Rotation

ecLoad

Apparent pointof rotation

(moves downwardas buttdisplacementincreases) / /

Conatantbutt slope

Load • Constant butt

V /21

-Apparent point of rotation

(moves upward asbutt slope increases)

(c) Illustration of Increase in Depth of (d) Illustration of Decrease in Depth ofApparent Point of Rotation Apparent Point of Rotation

I

t&

Soil failure

Kickout of shaft tip

Shaft failure orrigid body rotation

Lateral Load or Moment

(e) Typical Variation of Apparent Depth of Rotation with Load

Figure 59 - Analysis of Lateral Load Test

8

303

APPENDIX A

SUMMARY OF RESULTS OF INSTRUMENTED PILELOAD TESTS IN HONG KONG

304

305

CONTENTSPageNo.

TITLE PAGE 303

CONTENTS 305

A.I GENERAL 307

A.2 MOBILISED SKIN FRICTION ON PILES 307A.2.1 Replacement Piles 307A.2.2 Displacement Piles 308

A. 3 DATABASE OF INSTRUMENTED PILE LOAD TEST RESULTS 308

A.4 REFERENCES 308

LIST OF TABLES 311

LIST OF FIGURES 321

307

A.I GENERAL

This Appendix gives a summary of results of instrumented pile load tests in soils inHong Kong. The data were obtained from published papers and from local piling contractorsand consultants. Based on these data, the skin friction mobilised at the pile/soil interfaceduring pile load tests has been assessed and discussed below.

A.2 MOBILISED SKIN FRICTION ON PILES

A. 2.1 Replacement Piles

The mobilised skin friction values as determined from load tests on instrumented pilesare summarised in Tables Al and A2 for replacement piles and displacement pilesrespectively.

A number of tests on large-diameter bored piles and barrettes in Table Al indicatedthat the skin friction component is fully or substantially mobilised at a relative pile/soilsettlement of 1 % pile diameter.

The test results indicate a complex and erratic distribution of 'local' skin friction withdepth. Some of the results are known or suspected to have been a result of pile construction,e.g. filter cake problems, and cannot be taken to be representative of the behaviour of theground.

The average mobilised skin friction in saprolites have been plotted in Figures Al toA4 for replacement piles. Different symbols have been used in the Figures to delineate thequality of the data, which is described below.

In Figures Al to A4, results of loading tests for which the skin friction are fully orsubstantially mobilised are plotted as solid circles. In cases where the interpreted maximumskin friction is not substantially mobilised, hyperbolic curve fitting was used together withthe skin friction and shaft movement data to estimate the skin friction at 1% diameter shaftmovement. The measured maximum skin friction is shown as open square and theextrapolated skin friction at 1 % diameter shaft movement is denoted as solid square.

The test results derived from two bored piles C8-6-4 in Site 1 and TP1 and TP2 inSite 6 were suspected to have been affected by construction problems and may not berepresentative. The results are shown as open circle in the Figures.

For the skin friction values reported by Fraser & Kwok (1986), Davies & Chan (1981)and Evans et al. (1982), information regarding the shaft movement is not available.Therefore, the degree of mobilisation of skin friction is not known. These skin frictionvalues are represented by open triangles in Figure Al through A4.

The test results reported by Sayer & Leung (1987) have not been included in FiguresAl through A4 because the SPT N values of the saprolite at each caisson were not known.

It can be seen in the Figures that there is considerable scatter in the test results. The

308

variability may be related to the different methods of construction and workmanship, and theheterogeneous nature of the saprolites with intrinsic weak bonding which may be susceptibleto influence of pile construction (e.g. from stress relief and mechanical remoulding).However, it is noteworthy that the scatter of the results, although considerable, is comparableto that for load tests carried out in granular soils as reported by Meyerhof (1976) and Wright& Reese (1979).

A.2.2 Displacement Piles

The results of instrumented load tests on displacement piles are shown in Figures A5to A8. The symbols used are the same as for the replacement piles. For displacement piles,the relative movements required to fully mobilise the skin friction range typically from 5 mmto 15 mm (say about 1 to 3% of pile diameter).

In a number of the tests, the skin friction was not fully mobilised due to insufficientsettlement. No extrapolation of the data to ultimate skin friction was made in view of thefindings of Yiu & Lam (1990) which showed the danger of extrapolation of test results fordriven piles (see also Section 5.4.2). In addition, it should be noted that the post-peak dropin the strength along the interface between the pile and a bonded material can be significant(Coop & McAuley, 1992). Such strain-softening characteristics, particularly in the case oflong piles, will lead to a lower average mobilised strength. This type of behaviour can beassessed within the framework proposed by Murff (1980) or Randolph (1983). However, toquantify the effects, good quality tests would be required on the interface behaviour, such asdirect shear tests of the interface under constant normal stiffness conditions (Coop &McAuley, 1992).

A.3 DATABASE OF INSTRUMENTED PILE LOAD TEST RESULTS

The use of rational methods to back-analyse results of pile load tests on instrumentedpiles will lead to a better understanding of pile behaviour. However, it is evident that theamount of such pile load test data is limited. Therefore, the Geotechnical Engineering Officehas established a data-base of instrumented pile load test results and regularly updatesnormalising plots, such as those in Figures Al to A8, when more test results becomeavailable.

To facilitate access to pile load test data by all interested parties, practitioners areencouraged to submit such data to the Geotechnical Information Unit of the Civil EngineeringLibrary.

A.4 REFERENCES

Coop, M.R. & McAuley, J. (1992). The shaft friction of piles in carbonate soils.Proceedings of the International Conference on Offshore Site Investigation andFoundation Behaviour, London, pp 645-660.

309

Meyerhof, G.G. (1976). Bearing capacity and settlements of piled foundations. Journal ofGeotechnical Engineering, American Society of Civil Engineers, vol. 102, pp 197-228.

Murff, J.D. (1980). Pile capacity in a softening soil. International Journal of Numericaland Analytical Methods in Geomechanics, vol. 4, pp 185-189.

Randolph, MLF. (1983). Design considerations for offshore piles. Proceedings of theConference on Geotechnical Practice in Offshore Engineering, Austin, USA,pp 422-439.

Vesic, A.S. (1977). Design of Pile Foundations. National Co-operative Highway ResearchProgram, Synthesis of Highway Practice No. 42, Transportation Research Board,National Research Council, Washington D.C., 41 p.

Wright, SJ. & Reese, L.C. (1979). Design of large diameter bored piles. GroundEngineering, vol. 12, no. 8, pp 17-51.

Yiu, T.M. & Lam, S.C. (1990). Ultimate load testing of driven piles in meta-sedimentarydecomposed rocks. Proceedings of the Conference on Deep Foundation Practice,Singapore, pp 293-300.

U)J—<*o

311

LIST OF TABLES

Table PageNo. No.

Al Interpreted Skin Friction in Load Tests on Instrumented 313Replacement Piles in Hong Kong

A2 Interpreted Skin Friction in Load Tests on Instrumented 318Displacement Piles in Hong Kong

<Nr-Hm

Table Al - Interpreted Skin Friction in Load Tests on Instrumented Replacement Piles in Hong Kong (Sheet 1 of 5)

References

Holtet al(1982)

Linney(1983)

Ho (1992)

PileLength

(m)

16 0

36.5

32.8

36.8

PileDimension

(m)

1 n

1.0

1.2

1.2

Pile Construction

Bored pile -reverse circulationdrill with waterflush

Bored pile -constructionmethod not given

Bored pile -constructed byhammer, grab &casing under water- PP/F14

Ditto -Pile 14F8B

Stratum

Fill

Marinedeposit +Alluvium

Weatheredgranite

Fill +marine sand& clay

Alluvialsand

Weatheredgranite

Weatheredvolcanics

WeatheredVolcanics

MaximumMobilised

Average SkinFriction, rmax

(kPa)

31*

32*

129*

35*

42*

98*

30*

25

RelativePile/Soil

Settlement(5) at rniax

(mm)

6

5

39

10

29

23

3

5

Tl%D

31

28

88.4

35

39

91.5

30

50

Mean SPTValue, N

?

7

> 100

7

7

9

35

78

MeanOv

(kPa)

83

175

267.5

54

140

251

194.2

205.2

rnm/N

(kPa)

7

7

1.3

7

7

9

0.86

0.32

WN(kPa)

7

7

0.88

7

7

9

0.86

0.64

0max

( = rnm/O

0.37

0.18

0.48

0.65

0.30

0.39

0.15

0.12

M l ^ D

( = r,%D/O

0.37

0.16

0.33

0.65

0.28

0.36

0.15

0.24

FiguresAl to

A4

PI

P3

Pl l

P12

Table Al - Interpreted Skin Friction in Load Tests on Instrumented Replacement Piles in Hong Kong (Sheet 2 of 5)

References

Fraser &Kwok(1986)

)avies &Chan(1981)

PileLength

(m)

30

22.6

9

PileDimension

(m)

1 <

1.5

1 C

9

Pile Construction

Bored pileconstructed usinghammer grab &casing under water(except for bottom5 m where a reversecirculation drill wasused) - Pile 72/2

Bored pileconstructed usinghammer grab &casing under waterwith a concrete plugat the pile base -Pile 86/1

Bored pileconstructed usinghammer grab &casing under water -Pile 99/2

Bored piles

Stratum

Alluvium +2 m ofWeatheredgranite

Weatheredgranite

Alluvium

Weatheredgranite

Alluvium

Weatheredgranite

Weatheredgranite

MaximumMobilised

Average SkinFriction, 7max

(kPa)

26

21.5

16

80

8

23

50

RelativePile/Soil

Settlement

(mm)

9

9

9

9

9

9

9

Tl%D

9

9

9

9

9

9

9

Mean SPTValue,

N

9

55

15

80

28

65

42

Mean

(kPa)

63.3

184

48

133.7

38.4

120.1

9

- ,£,

(kPa)

9

0.39

1.1

1.0

0.29

0.35

1.2

- ,^j

(kPa)

9

9

9

9

9

9

9

Q

(-rmax/0

0.41

0.12

0.33

0.60

0.21

0.19

9

Q

(=WO

9

9

9

9

9

9

9

Mark inFiguresAl to

A4

P8

P9

P10

P16

Table Al - Interpreted Skin Friction in Load Tests on Instrumented Replacement Piles in Hong Kong (Sheet 3 of 5)

References

Sweeney &Ho (1982)

Sayer &Leung(1987)

Evans et al(1982)

Malone etal (1992)

Pratt(1989)

PileLength"(m)'

39

9

11.5

14

13.2

36

56.0

PileDimension

(m)

1.0

2.1

1.2

1.3

1.3

0.6 x 2.2

0.8 x 2.2

Pile Construction

Hand-dug caisson -jacking tests oncaisson rings

Hand-dug caisson -jacking tests oncaisson rings

Hand -dug caisson(timber stakesdriven ahead forstability) - P45

Hand-dug caisson(timber stakes

stability ) - P54

Ditto - Pile P141

Barrette constructedusing rectangulargrabs underbentonite

Barrette constructedusing rectangulargrabs underbentonite

Stratum

Weatheredgranite

Weatheredgranite

Fill/Alluvium/Weatheredgranite

Alluvium +Weatheredgranite

Weatheredgranite

Alluvium

Weatheredgranite

Weatheredgranite

Weatheredgranite

MaximumMobilised

Average SkinFriction, rnm

(kPa)

235*

70 - 100

130 - 170

34

18

27

58

52

126.7*

152*

RelativePile/Soil

Settlement(«) at rmax

(mm)

22

3 to 12

1 to 11

9

9

9

9

9

13

33

r,»

142.7

79.2-145.4

151 -213.5

9

9

9

9

9

128

120

Mean SPTValue,

N

200

140 (?)

200 (?)

27

19

43

27.5

60

132

65

Mean

(kPa)

665

9

9

142

86.9

126.3

49.5

253.4

276.0

370

_ _

(kPa)

1.2

9

9

1.26

0.95

0.63

2.1

0.87

0.96

2.3

_

(kPa)

0.71

9

9

9

9

9

9

9

0.97

1.9

(~TmJ O

0.35

9

9

0.24

0.21

0.21

1.2

0.21

0.46

0.41

_Pl*D

0.21

9

9

9

9

9

9

9

0.46

0.32

Mark irFiguresAl to

A4

C3

Cl

02

B3

B2

Table Al • - Interpreted Skin Friction in Load Tests on Instrumented Replacement Piles in Hong Kong (Sheet 4 of 5)

References

Site 1

Site 2

PileLength

(m)

49.3

52.1

40.6

42.2

48.2

65

f-l C

PileDimension

(m)

1.5

1.5

1.5

1.5

1.5

1.0

1 A

Pile Construction

Bored pile -constructed byhammer, grab &casing under water -PileC8-6-4

Ditto - Pile C8-7-1

Ditto - Pile C8-17-3

Ditto - Pile C8-17-4

Bored pileconstructed usinghammer grab &casing under water -WP13

Bored pile - reversecirculation drillunder bentonite -TP1

Ditto - TP2

Stratum

Weatheredgranite

Weatheredgranite

Weatheredgranite

Weatheredgranite

Weatheredgranite

Fill/Alluvium

Colluvium

Colluvium/ResidualSoil/Weatheredgranite

Weatheredgranite

Fill/Colluvium/Residual soil

Weatheredgranite

MaximumMobilised

Average SkinFriction, rmax

54*#

36

58

87

45.3

46*

48*

55*

155*

161

72

RelativePile/Soil

Settlement(5)at7n]ax

32

8

4

10

«1

1.6

7.2

2.2

3

7

6

T\%D

46.5

43.1

77.1

95.7

9

46

48

50

152

164

104

Mean SPTValue, N

106

80

107

65

104

21

18

41

92

26

68

Mean

(kPa)

290

360

302

270

318.6

108.3

268.5

451

623.5

211

627.2

(kPa)

0.51

0.45

0.54

1.3

0.44

2.2

2.7

1.3

1.7

6.2

1.1

(kPa)

0.44

0.54

0.72

1.5

9

2.2

2.7

1.2

1.7

6.3

1.5

(=T!:JO

0.19

0.10

0.19

0.32

0.14

0.42

0.18

0.12

0.25

0.58

0.11

<-£-o

0.16

0.12

0.26

0.35

9

0.42

0.18

0.11

0.24

0.59

0.17

Mark inFiguresAl to

A4

P4

P5

P6

P7

P13

P14

P15

Table Al - Interpreted Skin Friction in Load Tests on Instrumented Replacement Piles in Hong Kong (Sheet 5 of 5)

References

Site 4

Site 5

Site 6

Site 7

Site 8

PileLength

(m)

•40

48; testsection

5.2 m atbase

42.6

59.1

56.8

53

PileDimension

(m)

0.8 x 2.2

1.0

1.5

1.5

0.8 X 2.2

0.6 x 2.2

Pile Construction

Barrette constructedusing rectangulargrabs underbentonite

Bored pileconstructed usinghammer grab &casing under water

Bored pile - reversecirculation drillunder bentonite -TP1

Ditto - TP2

Barrette constructedusing rectangulargrabs underbentonite

Barrette constructedusing rectangulargrabs underbentonite

Stratum

Weatheredgranite

Weatheredgranite

Weatheredgranite

Weatheredgranite

Alluvium

Weatheredgranite

Weatheredgranite

MaximumMobilised

Average SkinFriction, rmax

(kPa)

104*

77* +

19*#

28*#82*#

94*

89*

51

RelativePile/Soil

Settlement

(mm)

18

10

19

1820

21

17

8

r l fD

95

77

19

2770.7

76

82

56

Mean SPTValue, N

80

140

97

77200

14

61

66

Mean

(kPa)

281.3

397.5

250

222.5456.5

248

410

328.1

(kPa)

1.3

0.55

0.20

0.360.41

6.7

1.5

0.77

(kPa)

1.2

0.55

0,20

0.350.35

5.4

1.3

0.85

(-£/*,

0.37

0.19

0.08

0.130.18

0.38

0.22

0.16

Pl^D

0.34

0.19

0.08

0.120.15

0.31

0.20

0.17

FiguresAl to

A4

B4

P2

P17

P18P19

B5

Bl

Notes : (1) * denotes substantially mobilised shear stress. (3) + denotes erratic strain gauge data.(2) / ? denotes information not available. (4) # denotes construction problems.

Table A2 - Interpreted Skin Friction in Load Tests on Instrumented Displacement Piles in Hong Kong (Sheet 1 of 2)

References

Premchittetal(1994)

PileLength

(m)

A1*! £i42.6

43.8

•

PileDimension

(m)

O c.5

0.5

Pile Construction

Precast prestressedconcrete pile - Pile118

Precast prestressedconcrete pile - Pile58

Stratum

Fill +MarineDeposits(silt)Marine Clay+ alluvialsandAlluvium(sand -fclay)AlluvialsandWeatheredgraniteFillMarine ClayMarine Clay4- AlluvialSandsAlluvialsandsAlluvialsands +Weatheredgranite

MaximumMobilised

Average SkinFriction, rmax

(kPa) ^

no

57

101

52

116

111*88*

88

96

37

RelativePile/Soil

Settlement($) at Tmax

(mm)

15

9

5.5

3

1

12.56.5

3.5

2

0.5

T1%D

79

47

99

62

187

8383

96

134

93

Mean SPTValue,

N

15

9

20

20

22

1712

15

17

18

Mean<TV'

(kPa)

72.9

129

177

237

317

80.9146.5

187

242

322

rmax/N(kPa)

7.3

6.3

5.1

2.6

5.3

6.57.3

5.9

5.6

2.1

WN(kPa)

5.3

5.2

5.0

3.1

8.5

4.96.9

6.4

7.9

5.2

_0max_(^Tmax/O

1.5

0.44

0.57

0.22

0.37

1.40.60

0.47

0.40

0.11

_0,%D_( = r1%D/av')

1.1

0.36

0.56

0.26

0.59

1.00.57

0.51

0.55

0.29

Mark inFiguresAl to

A4

Dl

D2

oo

Table A2 - Interpreted Skin Friction in Load Tests on Instrumented Displacement Piles in Hong Kong (Sheet 2 of 2)

References

Lam et al(1994)

Ng(1989)

Davies &Chan(1981)

Lee&Lumb(1982)

Site 9

PileLength

(m)

50.7

40.4

29

29

7

29.6

21.7

PileDimension

(m)

0.36 x0.38

ditto

0.5

0.5

7

0.61

0.5

Pile Construction

Steel H-pile - PP1

Steel H-pile - PP2

Precastprestressedconcrete pile -Pile B29

Ditto - Pile B34

Driven cast-in-place piles

Steel tubular pile

Precastprestressedconcrete pile

Stratum

Fill/Alluvium

Alluvium

Completelyweatheredmeta-siltstone

Fill/Alluvium

AlluviumCompletelyweatheredmeta-siltstone

Weatheredgranite

Weatheredgranite

Weatheredgranite

Marine clay

Weatheredmeta-siltstoneAlluvium/weatheredgranite

MaximumMobilised

Average SkinFriction, rmax

(kPa)

64*

61*

45*

7*67

54.8

174

129

100*

32*

63.7

137

RelativePile/Soil

Settlement(6) at

Mobilisedrmax (mm)

1310

5

59

5

6

6

7

7

7

12

T]%D

6461

45

752.5

35.5

153.5

120

7

7

7

88.9

Mean SPTValue,

N

1834

36

1535

45

16

23

30

4

30

20

Meanffy'

(kPa)

53.1153.4

331.9

68.7143.6

295.1

142

146

7

163

239

125

rmax/N(kPa)

3.61.8

1.3

0.471.9

1.2

10.9

5.6

3.3

8.0

2.1

6.9

T|%D/N

(kPa)

3.61.8

1.3

0.471.5

0.79

9.6

5.2

7

7

7

4.4

_ftn«_

( = Tmax/O

1.20.40

0.14

0.100.47

0.19

1.2

0.88

7

0.20

0.27

1.1

PI%D( = T|%D/O

1.20.40

0.14

0.100.37

0.12

1.1

0.82

7

7

7

0.71

Mark inFiguresAl to

A4

D3

D4

D5

D6

D7

D8

D9

Notes : (1) * denotes substantially mobilised shear stress,(2) ? denotes information not available.

LO

Co

321

LIST OF FIGURES

Figure PageNo. No.

Al Relationship between Skin Friction and Mean 323SPT Value, N for Replacement Piles

A2 Relationship between Skin Friction and Mean 324Vertical Effective Stress for Replacement Piles

A3 Relationship between Normalised Skin Friction and 325Mean SPT Value, N for Replacement Piles

A4 Relationship between Normalised Skin Friction and 326N/av' for Replacement Piles

A5 Relationship between Skin Friction and Mean 327SPT Value, N for Displacement Piles

A6 Relationship between Skin Friction and Mean 328Vertical Effective Stress for Displacement Piles

A7 Relationship between Normalised Skin Friction and 329Mean SPT Value, N for Displacement Piles

A8 Relationship between Normalised Skin Friction and 330N/av' for Displacement Piles

322

300

20 40 60 80 100 120 140

Mean SPT Value, N

160 180 200 220

Legend :

oA

Substantially mobilizedAffected by construction problemsDegree of mobilization unknown

Extrapolated skin friction at 1 % diameter movementMeasured maximum skin friction

Notes : (1) Possible problem with bentonite filter cake in P17, P18 & P19.(2) Construction problem in P4.(3) Erratic strain gauge data in P2,(4) All piles are in granitic saprolite except Pll & P12 which are in volcanic saprolite,(5) For details of pile construction, see Table Al.(6) Prefix of pile designation: B for barrette, C for hand dug caisson and P for bored pile.

Figure Al - Relationship between Skin Friction and Mean SPT Value, N for Replacement Piles

250

100 200 300 400 500

Mean Vertical Effective Stress, avf (kPa)

600 700

Legend ;

oA

Substantially mobilizedAffected by construction problemsDegree of mobilization unknown

Extrapolated skin friction at 1 % diameter movementMeasured maximum skin friction

Notes : (1) Possible problem with bentonite filter cake in P17, P18 & P19.(2) Construction problem in P4.(3) Erratic strain gauge data in P2.(4) All piles are in granitic saprolite except Pll & P12 which are in volcanic saprolite.(5) For details of pile construction, see Table Al.(6) Prefix of pile designation: B for barrette, C for hand dug caisson and P for bored pile.

Figure A2 - Relationship between Skin Friction and Mean Vertical Effective Stress for Replacement Piles

D

§ 4

O

'€*M 3

GO

I 2

'•0O

.22 1• *3c

055

~

• B2

^P14

P16 P7£| s84 0pi C3A * ' 1 •

P15a App — B3P110 c f S ,

A Bl f P4^Xp6 *P?Cl A . AP18J5p5 F3 P190

i i i i i i i i i i

0 20 40 60 80 100 120 140 160 180 200 220

Mean SPT Value, N

Legend :

Substantially mobilizedAffected by construction problemsDegree of mobilization unknown

Extrapolated skin friction at 1 % diameter movementMeasured maximum skin friction

Notes : (1) Possible problem with bentonite in filter cake P17, P18 & P19.(2) Construction problem in P4,(3) Erratic strain gauge data in P2.(4) All piles are in granitic saprolite except Pll & P12 which are in volcanic saprolite,(5) For details of pile construction, see Table Al.(6) Prefix of pile designation: B for barrette$ C for hand dug caisson and P for bored pile.

Figure A3 - Relationship between Normalised Skin Friction and Mean SPT Value, N for Replacement Piles

ie

o

GO"

^ 28

13

B2

P15 IB1 I

P55

B4

*»

°P19

QP17

to0\

0.1 0.2 0.3 0.4 0.5

N/avf (m2/kN)

0.6 0.7

Legend ;

oA

Substantially mobilizedAffected by construction problemsDegree of mobilization unknown

Extrapolated skin friction at 1 % diameter movementMeasured maximum skin friction

Notes : (1) Possible problem with bentonite filter cake in P17, P18 & P19.(2) Construction problem in P4.(3) Erratic strain gauge data in P2,(4) All piles are in granitic saprolite except PI 1 & P12 which are in volcanic saprolite.(5) For details of pile construction, see Table Al.(6) Prefix of pile designation: B for barrette, C for hand dug caisson and P for bored pile.

Figure A4 - Relationship between Normalised Skin Friction and N/OV' for Replacement Piles

300

250 -

_to

0 20 40 60 80 100 120 140 160 180 200 220

Mean SPT Value, N

Legend :

• Substantially mobilizedA Degree of mobilization unknown

Extrapolated skin friction at 1 % diameter movementMeasured maximum skin friction

Notes : (1) Piles D3 & D4 were only driven to length but not to set.(2) All piles are in granitic saprolite except D3, D4 & D8 which are in weathered meta-siltstone.(3) For details of pile construction, see Table A2.

Figure A5 - Relationship between Skin Friction and Mean SPT Value, N for Displacement Piles

250

100 200 300 400 500

Mean Vertical Effective Stress, avf (kPa)

LPt >00

600 700

Legend ;

• Substantially mobilizedA Degree of mobilization unknown

Extrapolated skin friction at 1 % diameter movementMeasured maximum skin friction

Notes : (1) Piles D3 & D4 were only driven to length but not to set.(2) All piles are in granitic saprolite except D3, D4 & D8 which are in weathered meta-siltstone.(3) For details of pile construction, see Table A2,

Figure A6 - Relationship between Skin Friction and Mean Vertical Effective Stress for Displacement Piles

ctf

12

10

I H 8

rf1.C/5 .

1 4

• D5

D9 1

iD6IDDl

AD7

AD8D2

D3QD4

vo

20 40 60 80 100 120 140

Mean SPT Value, N

160 180 200 220

Legend :

Substantially mobilizedDegree of mobilization unknown I.

Extrapolated skin friction at 1 % diameter movementMeasured maximum skin friction

Notes : (1) Piles D3 & D4 were only driven to length but not to set.(2) All piles are in granitic saprolite except D3, D4 & D8 which are in weathered meta-siltstone.(3) For details of pile construction, see Table A2.

Figure A7 - Relationship between Normalised Skin Friction and Mean SPT Value, N for Displacement Piles

12

10

I 8rfo

I .CO

I 2

• D5

CD

D2

D3

>D6

oD4

0.1 0.2 0.3 0.4

V' (m2/kN)

0.5 0.6 0.7

Legend :

• Substantially mobilizedA Degree of mobilization unknown

Extrapolated skin friction at 1 % diameter movementMeasured maximum skin friction

Notes : (1) Piles D3 & D4 were only driven to length but not to set.(2) All piles are in granitic saprolite except D3, D4 & D8 which are in weathered meta-siltstone,(3) For details of pile construction, see Table A2.

Figure A8 - Relationship between Normalised Skin Friction and N/avf for Displacement Piles

331

GLOSSARY OF SYMBOLS

332

333

GLOSSARY OF SYMBOLS

Ab cross-sectional area of pile base

Ac concrete cross-sectional area of pile

Acap area of pile cap

An cross-sectional area of pile element n

Ap cross-sectional area of pile

a exponent for stiffness efficiency factor

Be equivalent width of bell

b width of raft

C(m,t) compression of internal spring m at time t

Cd, Cs correction factors for depth and shape

CPT cone penetration test

CRP constant-rate-of-penetration (test)

c velocity of longitudinal stress wave through pile

c* elastic or recoverable movement of pile

c' apparent cohesion of soil or rock joint in terms of effective stress

cd spacing of discontinuities

cfa continuous-flight auger

cu undrained shear strength of soil

D pile width or width of foundation in the direction of rotation

D(m,t) displacement of pile element m at time t

Dj least lateral dimension of pile section

D2 diameter of pile, or diagonal width of pile where pile section is not circular

Db foundation base width or base diameter

334

Dc damping factor

Dr relative density of sand

Ds diameter of shaft

D'(m,t) plastic displacement of external spring m at time t

d depth factor

db depth below base of foundation

dc thickness of clay layer

dh height of fall of hammer

di thickness of soil layer i

dr foundation depth below rock surface

E pile driving energy

Eav weighted mean value of Young's modulus of founding material along lengthof pile

Ehf drained horizontal Young's modulus of soil

Ej modulus of soil layer i

En Young's modulus of pile element n

Ep Young's modulus of pile

Epe equivalent Young's modulus of pile

Er Young's modulus of rock

Es Young's modulus of soil

Evf drained vertical Young's modulus of soil

e eccentricity of load

ei free length of a free head pile above ground level or the equivalent freelength of a fixed head pile above point of virtual fixity

e1 effective eccentricity of load

F(m,t) force in internal spring m at time t

335

F5 deflection coefficient

Fb load at bottom of rock socket

Fd design safety factor on driving

FM moment coefficient

Fp force at a given pile section

Fs (global) factor of safety

Ft load at top of rock socket

Fv shear coefficient

f signal or excitation frequency

fb mobilisation factor for base resistance

fc specified grade strength of concrete

fm, ym multipliers to convert load and deflection of a single pile to a pile group

fn ultimate negative skin friction

fs mobilisation factor for skin friction

fy yield stress of steel

G shear modulus of soil

Gc characteristic shear modulus of soil

G'(m) quake for external spring m (or maximum elastic soil deformation)

g gravitational acceleration

H horizontal load on pile

Hg, Hp lateral load of a group pile and a single pile

Hu ultimate value of lateral load

Hx total applied horizontal load in x-direction

Hxi horizontal load on pile i

I influence factor for computing pile cap stiffness

336

In influence factor for estimating negative skin friction

Ip moment of inertia of pile

Ip*, Is* influence factors for pile settlement computation

Ix, ly moment of inertia of pile group with respect to x and y axes respectively

Ixy product of inertia of pile group about its centroid

Iyi moment of inertia of ith pile about its y-axis (orthogonal to the direction ofapplied force)

J(m) soil-damping constant at element m

jc damping coefficient in CASE analysis

js Smith damping coefficient

K pile stiffness factor

K(m) spring constant for internal spring m

Kc stiffness of pile cap

Kd dynamic stiffness of pile head

Kf overall foundation stiffness

Kg stiffness of pile group

Kh modulus of horizontal subgrade reaction of pile

K0 coefficient of earth pressure at rest

Kp coefficient of passive pressure

Kq, Kc passive pressure coefficients for short piles subject to lateral loading

Kr stiffness factor of rock socket under lateral loading

Ks coefficient of earth pressure

Ksp bearing pressure coefficient

Kv, KL pile stiffness under vertical and lateral loads

K'(m) spring constant for external spring m

337

k proportionality constant for the estimation of peak particle velocity due to piledriving

kh coefficient of horizontal subgrade reaction

ks coefficient of permeability of soil

L embedded length of pile

Lac active pile length

Lb foundation base length

Lc critical pile length

LJ length of pile element i

Lt the sound level that is exceeded for 1 % of the time interval considered

Lres resonating length

Ls length of rock socket

LW(A) a-weighted sound power level

1 length of raft

M applied bending moment on pile

Mf moment in a fixed-headed pile induced by a lateral force

MH, VH moment and shear force in pile due to applied horizontal force

Mmax maximum bending moment

M0 characteristic mobility

MQD Marble Quality Designation

MR Marble Rock Recovery Ratio

Mu ultimate moment of resistance of pile

Mx, My moment about centroid of pile group with respect to the x and y axesrespectively

Mx*, My* effective moment with respect to x and y axes respectively, taking intoaccount the symmetry of the pile layout

338

m pile element number

N uncorrected SPT blowcount

N mean SPT N value

N^ tan2 (45° + 4>'/2)

N! mean SPT N value corrected for an effective overburden pressure of 100 kPa

Nb number of blows of hammer per minute

Nc, Nq, N7 bearing capacity factors

Nf SPT blowcount after pile driving

Np GCO probe blowcount

NSF negative skin friction

Nu breakout factor

n number of observations or entities

nh constant of horizontal subgrade reaction

iip number of piles in a pile group

OCR overconsolidation ratio of soil

P applied vertical load

Pc load carried by pile cap

Pcr critical buckling load of pile

Pg load carried by pile group

Ph soil reaction per unit length of pile

Pi axial load on an individual pile i

PL concentrated horizontal force at pile tip due to passive soil resistance

PLJ50 point load index strength of a rock specimen of 50 mm diameter

Pm mobility at resonance (peak)

pn perimeter length of pile element n

339

Pn load due to ultimate negative skin friction

Pt, Pb applied load at pile head and pile base

p soil pressure

ppv peak particle velocity

pz unit passive resistance per unit width of pile at depth z

Qan allowable pile load

Qi5 Qi+1 load acting on top and bottom of pile element i

Qm mobility at anti-resonance (trough)

Qmax maximum test load

Qo ultimate concentric vertical load

Qs ultimate skin friction capacity under tension

QT total vertical load acting at the centroid of the pile group

Quit ultimate load capacity or ultimate resistance below the neutral point whenconsidering negative skin friction

Qv vertical component of the ultimate eccentric and inclined load

Qwt working load under tension loading

q bearing pressure on the rock mass

qa allowable bearing pressure

qb ultimate end bearing stress

R characteristic length of pile or stiffness factor = VEpIp/khD

R(m,t) force exerted by external spring m on element m at time t

RA ratio of pile cross-sectional area to area bounded by outer circumference ofpile

Rd dynamic component of pile penetration resistance or driving resistance

Rd(m) dynamic resistance of pile element m

R stiffness efficiency factor which is an inverse of the group settlement ratio

340

Rgs group settlement ratio of pile

Rh group lateral deflection ratio

Rn reduction factor for nh

RQD Rock Quality Designation

RS static component of pile penetration resistance

Rsu(m) ultimate static resistance of external soil spring m

rb empirical reduction factor for base resistance .

rc equivalent radius of pile cap for each pile

re reduction factor for load eccentricity

rf reduction factor for ultimate bearing capacity of vertical piles under eccentricand inclined loads

TI reduction factor for inclination of load

rm radius of influence of pile under axial loading

r0 pile radius or radius of an equivalent circular pile

s* permanent set of pile

sp centre-to-centre spacing of pile

T characteristic length of pile or stiffness factor = 5v/EpIp/nh

t time

V(m,t) velocity of pile element m at time t

Vi second peak in velocity upon arrival of reflected wave at pile top

Vb velocity of pile tip

Vc wave velocity in concrete

V0 first peak in velocity after falling mass contacts pile top

Vp velocity of pile at each segment

Vt pile head velocity

341

v particle velocity

W weight of ram

W(m) weight of element m

Wh weight of hammer

WL design working load of pile

Wpf effective self weight of pile

W effective self weight of the soil above the founding level

x distance between point of rotation and ground surface

xi? YI distance of pile i from y and x axes respectively

x mean sound level in dB(A)

Z pile impedance

Z l9 Z2 pile impedance below and above a given level where there is a significantchange in impedance

z depth below ground surface

zf vertical distance between point of virtual fixity and ground surface

a. adhesion factor

ab base interaction factor for a pile at a distance sp from a loaded pile

acp average pile interaction factor between pile and pile raft

a.L angle of inclination of applied load

as shaft interaction factor for a pile at a distance sp from a loaded pile

a' interaction factor for deflection of pile

/? shaft friction coefficient

j8*L, 7?*L dimensional length of pile for estimation of pile lateral deformation

01%D ratio of r1%D/avf

j8z damage classification factor = ratio of impedance of the pile section aboveand below a given level

342

|8' angle of inclination of pile

7rf effective unit weight of rock mass

7sf effective unit weight of soil

6 relative pile/soil settlement or pile settlement

<59o%Q Pile head settlement at 90% of failure or maximum test load

6b pile base movement

6d aperture of discontinuities

6das elastic deformation of pile element

Af frequency interval

Ah horizontal distance from pile axis

<5h lateral deflection of pile

5hg> ^hP lateral deflection of a pile group and a single pile

6j movement at the middle of pile element i

6M, MM, VM lateral pile movement, moment and shear force in pile due to applied moment

5max maximum pile head settlement

SQ pile head settlement at failure or maximum test load

<5res residual (or permanent) pile head settlement upon unloading from maximumtest load

6S angle of friction at pile/soil interface

At time interval

6t pile head settlement

5V vertical deformation of pile

f measure of radius of influence of pile

f] group reduction or efficiency factor

% efficiency of hammer (allowing for energy loss on impact)

343

*?r ratio of underream for underream piles

6 pile rotation at ground surface, or butt slope

0C constant butt slope

9B slope angle

X pile stiffness ratio

jLie microstrain

j>p Poisson's ratio of pile

vr Poisson's ratio of rock

*>s Poisson's ratio of soil

p Rate of variation of shear modulus of soil with depth

pc density of concrete

PC' degree of soil homogeneity over critical length Lc

crc uniaxial compressive strength of rock

crv' vertical effective stress

an_! standard deviation

<rv' mean vertical effective stress

TJ shear stress on pile element i

rs ultimate shaft (or skin) friction

?1%D average skin friction mobilized at relative settlement of 1% pile diameteiY

rmax maximum mobilised average skin friction

$ interaction factor for settlement analysis of a pile group

0cvf critical state friction angle of soil

<£/ residual angle of shearing resistance of soil

<£' angle of shearing resistance of founding material

\l/ angle of dilation of soil

344

angle of departure that the pile makes with the direction of loading

345

GLOSSARY OF TERMS

346

347

GLOSSARY OF TERMS

Barrettes. A variant of the traditional bored pile with rectangular cross-section. Therectangular holes are excavated with the use of grabs.

Base resistance. Load carrying capacity of pile due to bearing capacity of the soil below piletip.

Best-estimate parameter. Value of parameter which is representative of the properties ofmaterial in the field.

Composite piles. Special piles of various combinations of materials in driven piles.orcombinations of bored piles with driven piles.

Continuous-flight auger (cfa) piles. A proprietary piling system in which the bore is formedusing a flight auger and concrete or grout is pumped in through the hollow stem.

Driven cast-in-place piles. Piles formed by driving a steel tube into the ground to therequired set or depth and withdrawing the tube after concrete placement.

Hand-dug caisson. A bored pile in which the bore is formed manually by using hand toolsin stages.

Large-diameter bored piles. Bored piles of diameter greater than about 600 mm,e.g. machine bored piles.

Large-displacement piles. All solid driven piles, including precast concrete piles, and steelor concrete tubes closed at the lower end by a driving shoe or a plug.

Mini-piles. Small diameter piles which are formed by small drilling rigs with the use ofdown-the-hole hammers, rotary or rotary percussive drills and are subsequentlygrouted.

Mobilisation factors. Factors applied to skin friction and base resistance to estimate theallowable capacity of pile, taking into account different amounts of movement tomobilise skin friction and base resistance.

Precast concrete piles. Reinforced concrete piles, with or without prestress, cast and thendriven into ground.

Replacement pile. Pile formed by machine boring, grabbing or hand digging.

Saprolite. Soil derived from insitu rock weathering, which retains evidence of the originalrock texture, fabric and structure.

Skin friction. Load-carrying capacity of pile due to soil resistance developed at pile/soilinterface in response to applied load.

Small-diameter bored piles. Bored piles of small diameter less than about 600 mm.

348

Small-displacement piles. Driven rolled steel sections such as H-piles and open-ended tubularpiles.

Special piles. Particular pile types or variants of existing pile types introduced to improveefficiency or overcome problems related to special ground conditions.

Steel H-piles. Piles of rolled steel section of H-shape in cross-section.

Steel tubular piles. Preformed hollow steel piles of circular section.

GEOTECHNICAL ENGINEERING OFFICE PUBLICATIONS

Geotechnical Manual for Slopes, 2nd edition (1984), 306 p.(Reprinted, 1994).

Guide to Retaining Wall Design, 2nd edition (1993), 268 p.(Reprinted, 1994).

Guide to Site Investigation (1987), 368 p. (Reprinted, 1993).

Geoguide 1

Geoguide 2

Guide to Rock and Soil Descriptions (1988), 195 p. (Reprinted, Geoguide 31994).

Guide to Cavern Engineering (1992), 159 p. (Reprinted, 1994). Geoguide 4

Guide to Slope Maintenance (1995), 91 p. (English Version). Geoguide 5

Layman's Guide to Slope Maintenance (1995), 56 p. (Bilingual).' 56

Model Specification for Prestressed Ground Anchors, 2ndedition (1989), 168 p.

Model Specification for Reinforced Fill Structures (1989),140 p.

Mid-levels Study : Report on Geology, Hydrology and SoilProperties (1982), 265 p. plus 54 drgs.

Prediction of Soil Suction for Slopes in Hong Kong, by M.G.Anderson (1984), 243 p.

(Superseded by GCO Publication No. 1/85)

(Superseded by Geospec 1)

HKS74(US$21.5)

HKS48(US$16.5)

HK$83(US$28)

HK$58(US$18)

HK$36(US$13.5)

HK$30(US$6.5)

HK$40(US$7.5)

Free

HK$25(US$5.5)

HK$25(US$5.5)

HK$200(US$34)

GCO Publication HK$50No. 1/84 (US$9)

GCO PublicationNo. 2/84

GCO PublicationNo. 3/84

Geospec 1

Geospec 2

Review of Superficial Deposits and Weathering in Hong Kong, GCO Publication HKS40by ID. Bennett (1984), 58 p. (Reprinted, 1993). No. 4/84 (US$8)

Review of Hong Kong Stratigraphy, by J.D. Bennett (1984), GCO Publication HK$2586 p. No. 5/84 (US$5.5)

Review of Tectonic History, Structure and Metamorphism of GCO Publication HKS20Hong Kong, by J.D. Bennett (1984), 63 p. No. 6/84 (US$5)

(Superseded by GCO Publication No. 1/88) GCO PublicationNo. 1/85

Groundwater Lowering by Horizontal Drains, by D.J.Craig & GCO Publication HK$74I. Gray (1985), 123 p. (Reprinted, 1990). No. 2/85 (US$12)

(Superseded by GEO Report No. 9) GCO PublicationNo. 1/88

Review of Design Methods for Excavations (1990), 193 p.(Reprinted, 1991).

Foundation Properties of Marble and Other Rocks in the YuenLong - Tuen Mun Area (1990), 117 p.

Review of Earthquake Data for the Hong Kong Region (1991),115 p.

Review of Granular and Geotextile Filters (1993), 141 p.

Pile Design and Construction (1996), 348 p.

Report on the Kwun Lung Lau Landslide of 23 July 1994,2 Volumes, 400 p. (English Version).

'

Report on the Fei Tsui Road Landslide of 1 3 August 1 995,2 Volumes, 81 p. (Bilingual).

GCO PublicationNo. 1/90

GCO PublicationNo. 2/90

GCO PublicationNo. 1/91

GEO PublicationNo. 1/93

GEO PublicationNo. 1/96

H H

Report on the Shum Wan Road Landslide of 13 August1995, 2 Volumes, 63 p. (Bilingual).

(Hong Kong) Rainfall and Landslides in 1984, by J. Premchitt GEO Report(1991), 91 p. plus 1 dig.'(Reprinted, 1995). No. 1

HK$40(US$12)

HK$58(US$10)

HK$42(US$11.5)

HK$32(US$19)

HK$48(US$10.5)

Free

Free

Free

HK$118(US$17.5)

(Hong Kong) Rainfall and Landslides in 1985, by J. Premchitt GEO Report HKS126(1991), 108 p. plus 1 drg. (Reprinted, 1995). No. 2 (US$20)

(Hong Kong) Rainfall and Landslides in 1986, by J. Premchitt GEO Report HKS126(1991), 113 p. plus 1 drg. (Reprinted, 1995). No. 3 (US$20)

Hong Kong Rainfall and Landslides in 1987, by J. Premchitt GEO Report HK$122(1991), 101 p. plus 1 drg. (Reprinted, 1995). No. 4 (US$19.5)

Hong Kong Rainfall and Landslides in 1988, by J. Premchitt GEO Report HK$106(1991), 64 p. plus 1 drg. (Reprinted, 1995). No. 5 (US$16)

Hong Kong Rainfall and Landslides in 1989, by K.L. Siu (1991), GEO Report HK$ 126114 p. plus 1 drg. (Reprinted, 1995). No. 6 (US$20)

Aggregate Properties of Some Hong Kong Rocks, by T.Y. GEO Report HK$120Irian, A. Cipullo, A.D. Burnett & J.M. Nash (1992), 212 p. No. 7 (US$19.5)(Reprinted, 1995).

Foundation Design of Caissons on Granitic and Volcanic Rocks, GEO Report HK$62by T.Y. Irfan & G.E. Powell (1991), 85 p. (Reprinted, 1995). No. 8 (US$10.5)

Bibliography on the Geology and Geotechnical Engineering of GEO ReportHong Kong to December 1991, by E.W. Brand (1992), 186 p. No. 9(Superseded by GEO Report No.39)

Bibliography on Settlements Caused by Tunnelling, by E.W. GEO Report HK$48Brand (1992), 50 p. (Reprinted, 1995). No. 10 (US$8.5)

Direct Shear Testing of a Hong Kong Soil under Various GEO Report HK$136Applied Matric Suctions, by J.K. Gan & D.G. Fredlund (1992), No. 11 (US$21.5)241 p. (Reprinted, 1995).

Rainstorm Runoff on Slopes, by J. Premchitt, T.S.K. Lam, J.M. GEO Report HKS121Shen and H.F. Lam (1992), 211 p. (Reprinted, 1995). No. 12 (US$19.5)

Mineralogical Assessment of Creep-type Instability at Two GEO Report HK$87Landslip Sites, by T.Y. Irfan (1992), 136 p. (Reprinted, 1995). No. 13 (US$15)

Hong Kong Rainfall and Landslides in 1990, by K.Y. Tang GEO Report HK$112(1992), 78 p. plus 1 drg. (Reprinted, 1995). No. 14 (US$17)

Assessment of Stability of Slopes Subjected to Blasting GEO Report HK$75Vibration, by H.N.' Wong & P.L.R. Pang (1992), 112 p. No. 15 (US$12)

(Reprinted, 1995).

Earthquake Resistance of Buildings and Marine Reclamation GEO Report HK$48Fills in Hong Kong, by W.K. Pun (1992), 48 p. (Reprinted, No. 16 (US$8.5)

1995).

Review of Dredging Practice in the Netherlands, by S.T. Gilbert GEO Report HK$76& P.W.T. To (1992), 112 p. (Reprinted, 1995). No. 17 (US$12)

Backfilled Mud Anchor Trials Feasibility Study, by C.K. Wong GEO Report HKS50& C.B.B. Thorley (1992), 55 p. (Reprinted, 1995). No. 18 (US$9)

A Review of the Phenomenon of Stress Rupture in HDPE GEO Report HK$56Geogrids, by G.D. Small & J.H. Greenwood (1993), 68 p. No. 19 (US$9.5)(Reprinted, 1995).

Hong Kong Rainfall and Landslides in 1991, by N.C. Evans GEO Report HKS111(1992), 76 p. plus 1 drg. (Reprinted, 1995). No. 20 (US$16.5)

Horizontal Subgrade Reaction for Cantilevered Retaining Wall GEO Report HK$44Analysis, by W.K. Pun & P.L.R. Pang (1993), 41 p. (Reprinted, No. 21 (US$8)1995).

Report on the Rainstorm of 8 May 1992, by N.C. Evans (1993), GEO Report HK$126109 p. plus 2 drgs. (Reprinted, 1995). No. 22 (US$20)

Effect of the Coarse Fractions on the Shear Strength of GEO Report HKS126Colluvium, by T.Y. Irfan & K.Y. Tang (1993), 223 p. No. 23 (US$20)(Reprinted, 1995).

The Use of PFA in Reclamation, by J. Premchitt & N.C. Evans GEO Report HK$52(1993), 59 p. (Reprinted, 1995). No. 24 (US$9)

Report on the Rainstorm of May 1982, by M.C. Tang (1993), GEO Report HK$135129 p. plus 1 drg. (Reprinted, 1995). No. 25 (US$21)

Report on the Rainstorm of August 1982, by R.R. Hudson GEO Report HK$118(1993), 93 p. plus 1 drg. (Reprinted, 1995). No. 26 (US$17.5)

Landslips Caused by the June 1983 Rainstorm, by E.B Choot GEO Report HK$83(1993), 124 p. (Reprinted, 1995). No. 27 (US$13)

Factors Affecting Sinkhole Formation, by Y.C. Chan (1994), GEO Report HK$4037 p. (Reprinted, 1995). No. 28 (US$7.5)

Classification and Zoning of Marble Sites, by Y.C. Chan (1994), GEO Report HK$4037 p. (Reprinted, 1995). No. 29 (US$7.5)

Hong Kong Seawall Design Study, by P.M. Aas & A. Engen GEO Report HK$68(1993), 94 p. (Reprinted, 1995). No. 30 (US$11)

Study of Old Masonry Retaining Walls in Hong Kong, by Y.C. GEO Report HK$130Chan (1996), 225p. No.31 (US$21)

ECarst Morphology for Foundation Design, by Y.C. Chan & GEO Report HKS118W.K. Pun (1994), 90 p. plus 1 drg. (Reprinted, 1995). No. 32 (US$17.5)

An Evaluation of the Suitability of Decomposed Granite as GEO Report HK$38Foundation Backfill for Gravity Seawalls in Hong Kong, by E.B. No. 33 (US$7)Choot (1993), 34 p. (Reprinted, 1995).

A Partial Factor Method for Reinforced Fill Slope Design, by GEO Report HK$50H.N. Wong (1993), 55 p. (Reprinted, 1995). No. 34 (US$9)

Hong Kong Rainfall and Landslides in 1992, by P.K.H. Chen GEO Report HK$167(1993), 201 p. plus 2 drgs. (Reprinted, 1995). No. 35 (US$25.5)

Methods of Test for Soils in Hong Kong for Civil Engineering GEO Report FreePurposes (Phase I Tests), by P.Y.M. Chen (1994), 91 p. No. 36

Creep, Stress Rupture and Hydrolysis of Polyester Reinforced GEO Report HKS38Geogrids, by J.H. Greenwood (1995), 67 p. No. 37 (US$7.5)

Skin Friction on Piles at the New Public Works Central GEO Report HK$97Laboratory, by J. Premchitt, I. Gray & K.K.S. Ho (1994), 158 p. No. 38 (US$16.5)(Reprinted, 1995).

Bibliography on the Geology and Geotechnical Engineering of GEO Report HKS118Hong Kong to May 1994, by E.W. Brand (1994), 202 p. No. 39 (US$19)(Reprinted, 1995).

Hydraulic Fill Performance in Hong Kong, by C.K. Shen & GEO Report HK$90K.M. Lee (1995), 199 p. No. 40 (US$16)

Mineralogy and Fabric Characterization and Classification of GEO Report HK$70Weathered Granitic Rocks in Hong Kong, by T.Y. Man (1996), No. 41 (US$13.5)158 p.

Performance of Horizontal Drains in Hong Kong, by R.P. GEO Report HK$65Martin, K.L. Siu&J. Premchitt (1995), 109 p. No. 42 (US$17.2)

Hong Kong Rainfall and Landslides in 1993, by W.L. Chan GEO Report HK$110(1995), 214 p. plus 1 drg. No. 43 (US$18.5)

General Report on Landslips on 5 November 1993 at Man-made GEO Report HK$64Features in Lantau, by H.N. Wong &K.K.S. Ho (1995), 78 p. No. 44 (US$17)plus 1 drg.

Gravity Retaining Walls Subject to Seismic Loading, by Y.S. GEO Report HK$40Au-Yeung&K.K.S.Ho(1995),63p. No. 45 (US$8)

Direct Shear and Triaxial Testing of a Hong Kong Soil under GEO ReportSaturated and Unsaturated Conditions, by J.K.M. Gan & D.G. No. 46Fredlund(1996),217p.

Stability of Submarine Slopes, by N.C. Evans (1995), 51 p.

Strength Development of High PFA Content Concrete, by W.C.Leung & W.L. Tse (1995), 84 p.

AAR Potential of Volcanic Rocks from Anderson Road Quarries,by W.C. Leung, W.L. Tse, C.S. Mok & S.T. Gilbert (1995), 78 p.

Bibliography on the Geology and Geotechnical Engineering ofHong Kong to March 1996, by E.W. Brand (1996), underpreparation.

Bibliography on Settlements Caused by Tunnelling to March1996, by E.W. Brand (1996), under preparation.

Investigation of Some Major Slope Failures between 1992 and1995, by Y.C. Chan, W.K. Pun, H.N. Wong, A.C.O. Li & K.C.Yeo (1996), under preparation.

Environmental Aspects of Using Fresh PFA as Fill inReclamation, by K.S. Ho & P.Y.M. Chen (1996), underpreparation.

Hong Kong Rainfall and Landslides in 1994, by W.L. Chan(1996), under preparation.

Conventional and CRS Rowe Cell Consolidation Test on SomeHong Kong Clays, by J. Premchitt, K.S. Ho & N.C. Evans(1996), under preparation.

Geotechnical Area Studies Programme - Hong Kong andKowloon (1987), 170 p. plus 4 maps.

Geotechnical Area Studies Programme - Central NewTerritories (1987), 165 p. plus 4 maps.

Geotechnical Area Studies Programme - West New Territories(1987), 155 p. plus 4 maps.

Geotechnical Area Studies Programme - North West NewTerritories (1987), 120 p. plus 3 maps.

GEO ReportNo. 47

GEO ReportNo. 48

GEO ReportNo. 49

GEO ReportNo. 50

GEO ReportNo. 51

GEO ReportNo. 52

GEO ReportNo. 53

GEO ReportNo. 54

GEO ReportNo. 55

GASP I

GASP II

GASP III

GASP IV

HK$65(US$12.5)

HK$46(US$8.5)

HKS60(US$10.5)

HK$58(US$10)

HKS240(US$40)

HK$150(US$25)

HK$150(US$25)

HK$150(US$25)

Geotechnical Area Studies Programme - North New Territories(1988), 134 p. plus 3 maps.

Geotechnical Area Studies Programme - North Lantau (1988),124 p. plus 3 maps.

Geotechnical Area Studies Programme - Clear Water Bay(1988), 144 p. plus 4 maps.

Geotechnical Area Studies Programme - North East NewTerritories (1988), 144 p. plus 4 maps.

Geotechnical Area Studies Programme - East New Territories(1988), 141 p. plus 4 maps.

Geotechnical Area Studies Programme - Islands (1988), 142 p.plus 4 maps.

Geotechnical Area Studies Programme - South Lantau (1988),148 p. plus 4 maps.

Geotechnical Area Studies Programme - Territory of-HongKong (1989), 346 p. plus 14 maps.

Geology of Sha Tin, by R. Addison (1986), 85 p.

Geology of Hong Kong Island and Kowloon, by P.J. Strange &R. Shaw (1986), 134 p.

Geology of the Western New Territories, by R.L. Langford,K.W. Lai, R.S. Arthurton & R. Shaw (1989), 140 p.

Geology of Sai Kung and Clear Water Bay by P.J. Strange, R.Shaw & R. Addison (1990), 111 p.

Geology of the North Eastern New Territories, underpreparation.

Geology of Lantau District by R.L. Langford, J.W.C. James, R.Shaw, S.D.G. Campbell, P.A. Kirk& R.J. Sewell (1995), 173 p.

Geology of Yuen Long by D.V. Frost (1992), 69 p.

GASPV

GASP VI

GASPVII

GASP VIII

GASP IX

GASPX

GASP XI

GASP XII

GeologicalMemoir No. 1

GeologicalMemoir No. 2

GeologicalMemoir No. 3

GeologicalMemoir No. 4

GeologicalMemoir No. 5

GeologicalMemoir No. 6

Sheet ReportNo. 1

HK$150(US$25)

HKS150(US$25)

HK$150(US$25)

HK$150(US$25)

HK$150(US$25)

HKS150(US$25)

HKS150(US$25)

HKS150(US$25)

HK$50(US$9)

HK$78(US$12.5)

HK$97(US$17)

HKS87(US$13)

HK$136(US$28.2)

Free

Geology of Chek Lap Kok by R.L. Langford (1994), 61p. Sheet ReportNo. 2

Free

Geology of Tsing Yi by R.J. Sewell & J.A. Fyfe (1995), 43p. Sheet Report FreeNo. 3

Geology of North Lantau Island and Ma Wan by R.J. Sewell Sheet Report Free& J.W.C. James (1995), 46p. No. 4

Geology of Ma On Shan by R.J. Sewell (1996), 45p. Sheet Report FreeNo. 5

San Tin: Solid and Superficial Geology (1:20 000 map) (1989), Map HGM 20, HK$7C1 map. Sheet 2

Sheung Shui : Solid and Superficial Geology (1:20 000 map) Map HGM 20, HK$7C(1992), 1 map. Sheet 3

Kat 0 Chau : Solid and Superficial Geology (1:20 000 map) Map HGM 20, HK$7C(1993), 1 map. Sheet 4

Tsing Shan (Castle Peak) : Solid and Superficial Geology Map HGM 20, HK$7C(1:20 000 map) (1988), 1 map. Sheet 5

Yuen Long : Solid and Superficial Geology (1:20000 map) Map HGM 20, HK$70(1988), 1 map. Sheet 6

Sha Tin: Solid and Superficial Geology (1:20 000 map) (1986), Map HGM 20, HK$701 map. Sheet 7

Sai Kung Peninsula : Solid and Superficial Geology (1:20 000 Map HGM 20, HKS70map) (1989), 1 map. Sheet 8

Tung Chung : Solid and Superficial Geology (1:20000 map) Map HGM 20, HK$70(1994), 1 map. Sheet 9

Silver Mine Bay :• Solid and Superficial Geology (1:20 000 map) Map HGM 20, HK$70(1992), 1 map. Sheet 10

Hong Kong and Kowloon : Solid and Superficial Geology Map HGM 20, HKS70(1:20 000 map) (1986), 1 map. Sheet 11

Clear Water Bay: Solid and Superficial Geology (1:20 000 map) Map HGM 20, HKS70(1989), 1 map. Sheet 12

Shek Pik : Solid and Superficial Geology (1:20000 map) Map HGM 20, HK$70(1995), 1 map. Sheet 13

Cheung Chau : Solid and Superficial Geology (1:20 000 map), Map HGM 20, HK$70(1995), 1 map. Sheet 14

Hong Kong South and Lamma Island : Solid and Superficial Map HGM 20, HK$70Geology (1:20 000 map) (1987), 1 map. Sheet 15

Waglan Island : Solid and Superficial Geology (1:20 000 map) Map HGM 20, HKS70(1989),! map. Sheet 16

San Tin : Solid Geology (1 : 20 000 map) (1994), 1 map. Map HGM20S HK$70

Lo Wu : Superficial Geology (1:5 000 map) (1990), 1 map. Map HGP 5 A, HK$80Sheet 2-NE-D

Lo Wu : Solid Geology (1:5 000 map) (1990), 1 map. Map HGP 5B, HKS80Sheet 2-NE-D

Deep Bay : Superficial Geology (1:5000 map) (1989), Map HGP 5A, HKS801 map. Sheet 2-SW-C

Deep Bay : Solid Geology (1:5 000 map) (1989), 1 map. Map HGP 5B, HK$80Sheet 2-SW-C

Shan Pui: Superficial Geology (1:5 000 map) (1989), 1 map. Map HGP 5A, HK$80Sheet 2-SW-D

Shan Pui: Solid Geology (1:5 000 map) (1989), 1 map. Map HGP 5B, HKS80Sheet 2-SW-D

Mai Po : Superficial Geology (1:5 000 map) (1990), 1 map. Map HGP 5 A, HKS80Sheet 2-SE-A

Mai Po: Solid Geology (1:5 000 map) (1990), 1 map. Map HGP 5B, HK$80Sheet 2-SE-A

Lok Ma Chau : Superficial Geology (1:5000 map) (1990), Map HGP 5A, HKS801 map. Sheet 2-SE-B

Lok Ma Chau : Solid Geology (1:5 000 map) (1990), 1 map. Map HGP 5B, HK$80Sheet 2-SE-B

Man Kam To : Superficial Geology (1:5000 map) (1990), Map HGP 5A, HKS80Imap. Sheet 3-NW-C

Man Kam To: Solid Geology (1:5 000 map) (1990), 1 map. Map HGP 5B, HK$80Sheet 3-NW-C

Tin Shui Wai : Superficial Geology (1:5000 map) (1989), Map HGP 5A, HK$801 map. Sheet 6-NW-A

Tin Shui Wai: Solid Geology (1:5 000 map) (1989), 1 map. Map HGP 5B, HKS80Sheet 6-NW-A

Yuen Long • Superficial Geology (1:5 000 map) (1989), 1 map. Map HOP 5 A, HK$80Sheet 6-NW-B

Yuen Long • Solid Geology (1:5 000 map) (1989), 1 map. Map HGP 5B, HK$80Sheet 6-NW-B

Hung Shui Kiu : Superficial Geology (1:5 000 map) (1989), Map HGP 5 A, HK$801 map. Sheet 6-NW-C

Hung Shui Kiu : Solid Geology (1:5 000 map) (1989), 1 map. Map HGP 5B, HKS80Sheet 6-NW-C

Muk Kiu Tau : Superficial Geology (1:5000 map) (1990), Map HGP 5A, HKS801 map. Sheet 6-NW-D

Muk Kiu Tau : Solid Geology (1:5 000 map) (1990), 1 map. Map HGP 5B, HK$80Sheet 6-NW-D

Tsuen Wan (Part): Solid & Superficial Geology (1:5 000 map) Map HGP 5, HK$80(1995), 1 map. Sheet 6-SE-D

Ma On Shan: Solid Geology (1:5 000 map) (1996), 1 map. Map HGP 5B, HKS80Sheet 7-NE-D,C(part)

Chek Lap Kok : Solid & Superficial Geology (1:5 000 map) Map HGP 5, HKS80(1993), 1 map. Sheet 9-NE-C/D

Tung Chung Wan : Solid & Superficial Geology (1:5 000 map) Map HGP 5, HKS80(1995), 1 map Sheet 9-SE-A

Yam 0 Wan : Solid & Superficial Geology (1:5 000 map) Map HGP 5, HK$80(1995), 1 map. Sheet 10-NW-B

SiuHo : Solid & Superficial Geology (1:5 000 map) (1994), Map HGP 5, HK$801 map. Sheet 10-NW-C

Chok KoWan (Penny's Bay) : Solid & Superficial Geology Map HGP 5, HKS80(1:5 000 map) (1994), 1 map. Sheet 10-NW-D

Ma Wan: Solid and Superficial Geology (1:5 000 map) (1994), Map HGP 5, HK$80Imap. : Sheet 10-NE-A

Tsing Yi: Solid & Superfical Geology (1:5 000 map) (1995), Map HGP 5, HK$80Imap. Sheet 10-NE-B/D

Pa Tau Kwu : Solid and Superficial Geology (1:5 000 map) Map HGP 5, HK$80(1994), Imap. Sheet 10-NE-C

Tai Ho : Solid and Superficial Geology (1:5 000 map) (1995), Map HGP 5, HK$801 map. Sheet 10-SW-A

ORDERING INFORMATION IS GIVEN ON THE NEXT PAGE

Copies of GEO publications (except Sheet Reports, 1:5 000 maps and other reports which are free of charge) may beordered by writing to :

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Publications (Sales) Office,Information Services Department,28th Floor, Siu On Centre,188 Lockhart Road, Wan Chai,Hong Kong.

The Information Sendees Department will issue an invoice upon receipt of a written order.• '

In Hong Kong, publications may be directly purchased from::

Government Publications Centre,Ground Floor, Low Block,Queensway Government Offices,66 Queensway,Hong Kong.

Requests for copies of Geological Survey Sheet Reports and other reports which are free of charge should be directed to :'

Chief Geotechnical Engineer/Special Projects, 1fWlfGeotechnical Engineering Office, ±?ftlCivil Engineering Department,Civil Engineering Building,101 Princess Margaret Road,Homantin, Kowloon,Hong Kong.

1:5 000 maps may be purchased from:1:5 000 :

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All prices given in this List are for information only and may be changed without notice. The US$ prices shown are foroverseas orders and are inclusive of surface postage to anywhere in the world. An additional bank charge of HKS50 orUS$6.50 is required per cheque made in currencies other than Hong Kong dollars. Cheques, bank drafts or moneyorders must be made payable to HONG KONG GOVERNMENT,

* >

XOblfi3Slb

HK 624.154 P62Pile design and constructionHong Kong : the Office, 1996.

NC T FOR N-