stadium southland roof collapse report · 2012. 5. 11. · joint australian and new zealand...

STADIUM SOUTHLAND ROOF COLLAPSE REPORT

© Structuresmith Ltd 2012 © Hyland Consultants Ltd 2012 2 9 May. 12

REPORT PREPARED BY:

ASHLEY SMITH

ME, MIPENZ, CPEng, IntPE(NZ)

DR CLARK HYLAND

PhD, BE(Civil), BCom, CWE-NZIW, DipCL CPEng

Hyland Fatigue + Earthquake Engineering

Report Issue Date: 11 May 2012

Report Issue Status: Final

Limitations and Applicability of Report:

This report has been prepared solely on the basis of instructions from the Department of Building and Housing (“DBH”) and in accordance with the guidelines and scope set by DBH. No liability is accepted by Hyland Consultants Limited (“HCL”) or Structuresmith Ltd (“SSL”) or agents of HCL or SSL in respect to its use for any other purpose or by any other person or organisation.

This disclaimer shall apply notwithstanding that the report may be made available to other persons to fulfil a legal requirement.

This report has not been prepared for the purpose of determining liability or fault or blame.

HCL and SSL accept no liability for use of this report other than for the purpose referred to above, nor does it accept liability for matters which fall outside the scope set by DBH.

HCL and SSL retain copyright of this report unless HCL and SSL have expressly agreed otherwise with the Client in writing. This report may be reproduced and published but only in full and provided that it is not purported to be published under authority from HCL and SSL and that, in every case, acknowledgement is made of its source and that this disclaimer is incorporated in its publication.

Any opinions of HCL and SSL contained in this report are for information purposes only. HCL and SSL do not accept liability to any party, including DBH, for any statement contained in any opinion.

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© Structuresmith Ltd 2012 © Hyland Consultants Ltd 2012 PAGE 5 9 May. 12

TABLE OF CONTENTS

Overview ..................................................................................................................... 9 Scope of the investigation ......................................................................................... 10 Timing of Report ........................................................................................................ 11 Building History ......................................................................................................... 11 Collapse Sequence ................................................................................................... 11 Causes of the Collapse ............................................................................................. 12

Heavy Snow Fall 13

Construction Defects 13

Design and Detailing Issues 14

Problems with Remedial Works 15

Compliance with Standards and Drawings ................................................................ 15 Recommendations..................................................................................................... 16

Objectives and Scope................................................................................................ 19

Information Gathering ................................................................................................ 21 Determination of Collapse Sequence ........................................................................ 21 Site Examination ....................................................................................................... 22 Laboratory Examination ............................................................................................. 22

Stadium Location ....................................................................................................... 23 Stadium Facilities ...................................................................................................... 23 Structural Description ................................................................................................ 24

Background ............................................................................................................... 35 Shop Roof Collapse................................................................................................... 35 Fly Brace Restraint Buckling in Light Industrial Building ............................................ 36

Introduction ................................................................................................................ 41 Methodology .............................................................................................................. 41 Applicable Structural Design Standards .................................................................... 42 Summary of Findings from the Structural Evaluation ................................................ 44

Community Courts Roof Trusses T1 to T9 44

South Wall Grid 5 Column Heads 44


<TABLE OF CONTENTS continued


Connection of the Spine Trusses to the Concrete Columns 44

Spine Truss Support Columns C13 to C16 45

Roof Purlins 46

Summary Recommendations .................................................................................... 55 Design of Buildings Subject to Snow Loading 55

Construction of Buildings Subject to Snow Loading 56

Detailed Recommendations ...................................................................................... 56 Joint Australian and New Zealand Loadings Standard AS/NZS1170 56

Steel Design and Construction Standards NZS 3404 57

Concrete Design and Construction Standards NZS 3101 and NZS 3109 58

New Zealand Building Code and Associated Compliance Documents 58

Design Review and Construction Monitoring 58

Introduction ................................................................................................................ 65 Design Snow Loads and Load Combinations ............................................................ 65 Structural 3D Computer Analysis .............................................................................. 68

Methodology .............................................................................................................. 71 Roof Purlins ............................................................................................................... 71

Design Basis Evaluation of the Purlins 71

Failure Basis Evaluation of the Purlins 74

Discussion of Purlin Performance 76

South Wall Column Heads Connections .................................................................... 77 Design Basis Evaluation of South Wall Column Heads Connections 77

Failure Basis Evaluation of South Wall Column Heads Connections 77

Discussion of South Wall Column Heads Connections Performance 78

Community Courts Roof Trusses .............................................................................. 78 Design Basis Evaluation of Community Courts Trusses T1 to T9 78

Failure Basis Evaluation of Community Courts Trusses 81

Connections of Community Court Trusses T1 to T9 at Grid 4 ................................. 105 Design Basis Evaluation of T1 to T9 Connections to Grid 4 105

Failure Basis Evaluation of T1 to T9 Connections to Grid 4 105

Connections of Trusses T10 and T11 to Columns C13 to C16 ............................... 108 Design Basis Evaluation of T10 and T11 to Columns C13 to C16 108

Failure Basis Evaluation of T10 and T11 to Columns C13 to C16 108

Discussion of T10 and T11 to Columns C13 to C16 Performance 109


<TABLE OF CONTENTS continued


Concrete Columns C13 to C16 ................................................................................ 112 Design Basis Evaluation of Columns C13 to C16 112

Failure Basis Evaluation for Concrete Columns C13 to C14 116

Discussion of Concrete Columns C13 and C14 Performance 116

Failure Basis Evaluation for Concrete Columns C15 to C16 116

Discussion of Concrete Columns C15 and C16 Performance 117

Discussion of Concrete Columns C13, C14, C15 and C16 Performance 117

Introduction .............................................................................................................. 119 Community Courts Roof Truss Steel Wall Thickness .............................................. 119 Weld Quality ............................................................................................................ 119 Complete Penetration Weld Thickness.................................................................... 119 Weld Metal Properties ............................................................................................. 121 Spine Truss to Column Bolt Properties.................................................................... 121 Community Truss to Spine Truss Threaded Rod Properties ................................... 122 Community Trusses Top and Bottom Chord Tensile Properties .............................. 122 Paint Cracking Strain ............................................................................................... 123 Reinforcing Steel Properties .................................................................................... 124 Truss Top Chord Strut Compression Behaviour ...................................................... 125 Column Concrete Compressive Strength and Modulus ........................................... 126

3D Model of Roof Structure ..................................................................................... 127 Trusses 1 to 9 Modification Drawing........................................................................ 128 Collapse Sequence Diagram ................................................................................... 129



EXECUTIVE SUMMARY

OVERVIEW

The roof of the Stadium Southland building collapsed at around noon on 18 September 2010 following a heavy fall of snow in the area. There were no injuries or loss of life.

The investigation has shown that the stadium roof collapsed due to a combination of factors that included the heavy snowfall, construction defects, design detailing issues and problems with design changes and remedial works undertaken during construction.

The investigation found that the snow loading alone would not have been sufficient to cause the collapse of the Stadium Southland roof if its trusses had been fabricated to comply with the standards of the day.

Defects in the construction of the steel tube truss roof, particularly those portions that involved welding on site, appear to have significantly reduced the ability of the roof to sustain the intended design loads and the snow load experienced at the time of the collapse.

Specific factors that contributed (or may have contributed) to the collapse of the stadium include:


EXECUTIVE SUMMARY continued


A heavy snowfall close to a 1 in 250 year snow design level that may have resulted in snow loads on the roof approaching but not exceeding the load requirements that were in place when the stadium was built (1999/2000)

Construction defects and vulnerabilities:

o substandard connection of some building elements (such as community court trusses to the spine trusses, and spine trusses to supporting columns)

o substandard steelwork fabrication, especially site welding which weakened the structure (including insufficient welds, no welding in certain areas, incomplete welds, welds applied to painted surfaces, unwelded packing plates and incorrect preparation of welds)

o some critical structural members were supplied with wall thicknesses less than specified (eg thinner truss member wall thicknesses for some community court trusses)

o remedial works that were not undertaken in total compliance with the drawings; and/or the Steel Structures Standard; and/or the Welding Standard.

o problems with installation of strengthening and splice packing plates on Community Courts trusses

Design Detailing Issues:

o site splices in the Community Courts trusses at the highly loaded mid-span location

o design modifications to the Community Court trusses that were difficult to execute at a late stage in construction and resulted in “simply” supported trusses over the large community courts area

o spine trusses supports that were laterally flexible and did not have the strength or bracing to resist the westward drift that occurred in the collapse

o connections of the Community Courts trusses to the south wall concrete columns that were fragile

o non-compliant steel shear reinforcement in the concrete columns (C13-C16) supporting the main roof trusses (T10 and T11) as required by the Concrete Structures Standard

Inspection and Construction Monitoring:

o Most of the defects found to have been material to the collapse could have been identified by normal inspection procedures required by the Steel Structures Standard.

SCOPE OF THE INVESTIGATION

The technical investigation into the stadium collapse was undertaken for the Department of Building and Housing by StructureSmith Limited (SSL)




and Hyland Consultants Limited (HCL). The investigation involved determining the snow loading, site visits, reviewing drawings and specifications, laboratory examination and testing of the building remnants, the use of various structural analyses to develop an understanding of the building response and collapse-critical components, and the formulation of recommendations.

TIMING OF REPORT

The reason the stadium investigation report was significantly delayed was due to the Christchurch earthquake on 22 February 2011 one of the authors of this report was caught up providing on-ground assistance in the immediate aftermath of the Canterbury Earthquake of 22 February and both were subsequently contracted by the Department of Building and Housing to investigate the collapse of the CTV Building that collapsed on 22 February 2011.

BUILDING HISTORY

Construction of the Stadium commenced in 1999 and the building was fast-tracked so that it could be completed within seven months. The building provided a 9108m2 indoor sports facility for the Southland region with multi-purpose Community Courts, events courts, a climbing wall, seating, full service amenities, lounge, bar, kitchen and off-street parking facilities. The design of the stadium featured clear spans over two large inter-linked spaces comprising the Community Courts area and the events court area.

During the construction of the stadium excessive deflections or sagging was observed in the roof trusses that had been erected above the Community Courts area. Modifications had to be made to the stadium at a late stage of construction to address a range of problems including the deflections, the design capacity of the Community Court trusses, the connections to the south wall columns, the connections of the Spine trusses to their supporting columns and cracking in the spine truss columns.

Remedial works were undertaken to address the structural issues identified in a 1999 Review Report prior to construction being completed. These remedial works would have been, in the opinion of the authors, complex and difficult to achieve without skilled and experienced steel construction personnel and appropriate supervision. The adequacy of those remedial works was material to the collapse that occurred on 18 September 2010.

COLLAPSE SEQUENCE

The main areas of the stadium that collapsed as seen in Figure 2 were:

part of the end walls on the eastern and western sides of the stadium

roof purlins and roof trusses above the Community Courts




some roof purlins and one end of the main roof trusses above the events court

the main roof ‘spine’ trusses and their western supporting columns located between the Community Courts and the Events Courts.

The only person reported to be in the building at the time of collapse was not in the area affected by collapse. Tennis players who had just left the facility noted hearing a loud ‘crack’ or explosive sound come from the building as it collapsed and said they saw the doors at the eastern end of the Community Courts blow open and a roof panel fly past.

The stadium collapse is likely to have been initiated by the compression peeling and crushing failure of the mid-span top chord in one of the roof trusses over the eastern end of the Community Courts. The collapse then progressed through other roof trusses, causing a westward displacement of the trusses and their supporting columns. As the roof displaced westward, the bottom chord bolts that were connecting the spine trusses to two columns fractured. The spine trusses then fell to the ground, collapsing other columns. At the western end of the Community Courts, one of the roof trusses fell to the floor soon after and in response to the collapse of the trusses at the eastern end. Welds failed and strengthening plates peeled away on some of the roof trusses.

CAUSES OF THE COLLAPSE

The stadium roof collapsed due to a combination of factors that included the heavy snowfall, construction defects, design detailing issues and problems with remedial works undertaken during construction. These factors are outlined below.

East

South

Community Courts

Events Courts

Spine Trusses




Heavy Snow Fall

The snow load on the roof at the time it collapsed has been calculated to be 0.30 kPa (based on testing top-chord compression capacities). NIWA measured the snow on ground load to be 0.45 kPa on the day after the collapse event. This is close to the 1 in 250 year snow on ground load of 0.51 kPa calculated using the loading standard AS/NZS 1170.3 :2003. The factored roof design snow load using that loading standard would have been 70% of the snow on ground load, or 0.36 kPa, but not less than a minimum requirement of 0.40 kPa. However the snow load on the roof at failure determined in Appendix A was not estimated using this snow load on roof to snow on ground ratio.

It appears that the snow load may have increased from the time of the roof collapse as sleet, rain and some snow continued to fall, as indicated by the collapse of a Shop Roof nearby the following day.

The loading requirement for the stadium, when it was originally designed in 1999, was for a factored roof design snow load of 0.40 kPa combined with a factored dead load or self-weight of 0.41 kPa, giving a total factored design load of 0.81 kPa. The actual self-weight of the roof was calculated to be 0.34 kPa at the time of the collapse, allowing snow loading of 0.47 kPa to occur before the factored design loading was exceeded. At that point less than a 1% probability of failure occurring would be expected based on the safety indices built into New Zealand loading and steel structures design standards.

Therefore the heavy snow load alone is not able to explain the roof collapse.

Construction Defects

Construction defects were found to be critical in weakening the structure that would have otherwise been able to support the snow loads of 18 September 2010. The design resulted in a fragile structure that was susceptible to progressive collapse due to:

Connection of Elements

The investigation identified defects in the Community Court trusses and their connections, as well as the connections of spine trusses T10 and T11 to their supporting columns (C13 and C14).

The connections of the Community Court trusses to the south wall concrete column heads were found to be brittle and were unable to prevent the trusses being pulled off their supports. Some of the bolts to the top of the columns appeared to be missing.

The connections of the Community Courts truss T1 to T10 were compromised by site welding of the bottom chord to the support cleat. This prevented movement and caused greater demand on the top chord connection.




The connection of spine trusses T10 and T11 to columns C13 and C14 was made with packing plates that were not welded in position which would have increased the demands on the bolts connecting the trusses to the concrete columns.

However threaded rods connecting the Community Courts truss T1 to spine truss T10 and the bolts connecting spine trusses T10 and T11 to column C13 to C16 were found in the Laboratory Examination to be significantly stronger than was specified. This may have prevented collapse occurring at these locations at lower snow loads than what the structure actually collapsed at.

Welding

A range of welding deficiencies were identified that weakened the stadium structure including:

o top chords of truss T1 to T5 (at the mid-splice locations) had insufficient end bearing and welding at the splice plates

o quarter point splices of trusses T4 and T5 had insufficient welding

o no welding of the top surfaces of the top chords of the Community Court trusses at most splice locations

o top chord side strengthening plates were not joined by complete penetration welds

o welds attaching the strengthening plates (to sides of top chords) were applied to painted rather than to cleaned surfaces

o bolted connection of spine truss T10 to column C14 and T11 to C13 had unwelded packing plates

o in all trusses examined the mid-span bottom chord site welded splice had been incorrectly prepared and welded

Installation of Strengthening Plates

Several problems were found with the installation of the strengthening plates including the top chord side strengthening plates not being installed continuously past chord splice locations and splices on trusses T4 and T5 not having end bearing splice plates installed.

Design and Detailing Issues

Other design factors that contributed to the collapse included:

site splices in the Community Courts trusses at the most highly-loaded mid-span location.

the spine truss supporting structure did not have the lateral strength or stiffness to resist the westward drift imposed on it as the Community Court trusses began to collapse




design modifications that were difficult to execute at a late stage in construction

the design modifications resulted in simply supported trusses over the Community Courts reducing redundancy and did not reduce the fragility of their connections to the south wall concrete columns

non-compliant steel shear reinforcement in the concrete columns (C13-C16) supporting main roof spine trusses T10 and T11, with the spacing of reinforcing ties less that required in the applicable Concrete Structures Standard (NZS 3101: 1995)

Problems with Remedial Works

Modifications were made to the stadium structure at a late stage of construction to address problems including: excessive deflections; design capacity of the Community Court trusses; south wall column head connections; spine truss connections to columns; and cracking of spine truss columns. The investigation found that portions of the on-site steel fabrication and welding for these remedial works were not compliant with the drawings, and/or the Steel Structures Standard; and/or the Welding Standard. The remedial works were considered, in the opinion of the authors, to have been complex and difficult to achieve without skilled and experienced steel construction personnel and appropriate supervision.

COMPLIANCE WITH STANDARDS AND DRAWINGS

The areas of non-compliance with building standards and drawings identified during the investigation included the following:

the on-site steel fabrication and welding for the remedial works were not found to be compliant with the Drawings, and/or the Steel Structures Standard; and/or the Welding Standard

some Community Courts truss members were found to have steel tube wall thicknesses less than had been specified. However this was not found to have initiated the collapse.

the spacing of reinforcing ties in the concrete columns was found to be less that required in the applicable Concrete Structures Standard (NZS3101: 1995)

most of the substandard steel fabrication could have been identified by normal inspection procedures required of welding supervisors and construction reviewers, as specified by the Steel Structures standard (NZS3404:1997). This aspect of construction was therefore non-compliant at some critical locations.




RECOMMENDATIONS

The authors recommend that:

1. Research be undertaken to improve understanding of snow loadings for New Zealand sub-alpine coastal areas, in particular the relationship between roof and ground snow loads due to the effect of roof slope.

2. Review the statistical reliability basis of snow loadings, return periods and corresponding load and materials strength safety factors for the loadings, and the steel and concrete structures standards.

3. Public buildings with long span roofs expected to be affected by snow loadings are not opened for public use until the adequacy of the construction of all roof structural components have been appropriately inspected and certified.

4. Snow overload alarms be installed in public buildings with long span roofs subject to snow to warn occupants to exit the buildings when snow loads exceed the specified design limits.

5. Guidelines be developed for the design of roof structures subject to snow loadings to prevent progressive collapse in the event of snow overload. Require the connections of roof rafters and trusses to supports, to be sufficient to prevent them being pulled from their supports should those members suffer overload failure.

6. Require that the columns and walls supporting roof structures subject to snow loading be sufficiently stiff, strong or braced to resist transverse drifts imposed by collapsing rafters under overload conditions, thereby preventing progressive collapse.

7. Mandatory minimum levels of competency be set for companies and key personnel undertaking the construction of steel roof structures of public buildings subject to snow loadings.

8. Mandatory levels be set for independent third party design review and construction monitoring by appropriately qualified people for steel roof structures of public buildings subject to snow loadings.

9. Structural engineers be required to identify on the drawings collapse critical components of important roof structures subject to snow loading, for particular attention by reviewers, constructors and construction monitors.




10. Review the adequacy of the extent of concrete column buckling provisions in the Concrete Structures Standard NZS 3101.

More detailed recommendations are also provided in the body of the investigation report covering: general structural design provisions; snow load allowance; steel design and construction standards; concrete design and construction standards; and building consent and approval processes.

Disclaimer: This Executive Summary summarises the key points of this report and is not intended to be a substitute for the report in its entirety. The Executive Summary should be read in conjunction with the whole report and the reader should not act in reliance of the Executive Summary alone.




Purlins connected to T4 broke away from T3 from Grid 5 to midspan of T3

Top chord compression failure of T4 and T5 at quarter point splice That

had no end bearing plate

Spine trusses T10 and T11 bottom chord Connection 2-TCM 24 to Columns C13

and C14 fracture in combined tension and

shear

1500 mm vertical displacement of T1 after top chord compression failure at

splices of T1 to T5

Spine trusses T10

and T11 displaced

990 mm westward after collapse

T1 to T5 broke free from T10 upon impact with floor, deforming tie vertical struts in T10

Flexural compression hinging of columns C15 and C16 at underside of

T10 and T11 and at Control Room floor

Trusses T4 and T5 were pulled westwards off columns at Grid 5

Roof purlins and diagonal roof plane

bracing pulled collapsing Trusses T1

to T5 and roof westwards

T9 was pulled eastwards off column

support on Grid 5

T9 collapsed as bottom chord strengthening plates yield and

welds fracture

Likely tension failure of top chord splice from uplift of T9 due to

internal bursting pressure from collapse of T1 to T5, T10 and T11

C11

C10

C9

C8

C7

C6

C5

C4

C3



1 INTRODUCTION

OBJECTIVES AND SCOPE

The Department of Building and Housing (“the DBH”) appointed the authors to prepare an independent structural report to identify the causes of the Stadium Southland (“the Stadium”) roof collapse. DBH also instructed that recommendations be made as to whether any changes to the Building Code and related documents were required. The authors acknowledge that this report will be used by DBH to determine what action, if any, it needs to help prevent this type of event from occurring in the future.

The DBH set the following requirements for the report:

Determine the snow loadings required by NZS 4203:1992 and AS/NZS 1170.3:2003, B1 with Amendment 9. Also consider the site specific assessment of the actual snow on ground load reported by NIWA on the day after the collapse.

Visit the site to examine the collapsed structure.

Review available Drawings and specifications of the structure.

Carry out structural analysis as required to assess actions, and calculate capacities of collapse critical components of the Stadium.

Determine the likely sequence and reasons for the collapse.

Arrange and report on the laboratory examination and testing of selected salvaged structural components.

Make recommendations for improvements to New Zealand design and construction practices and standards, and the understanding of snowfall hazards as they relate to structural performance and to the New Zealand Building Code and related documents as required.



2 REPORT METHODOLOGY

The development of this report into the collapse of the Stadium Southland roof included:

INFORMATION GATHERING

The following documents were made available to the authors:

Summary listing of consents for the property

Building consent, remedial strengthening work and Construction Issue drawings (“the Drawings”)

A structural review report dated December 1999, undertaken during the construction of the Stadium (‘the 1999 Review Report”)

Section 31 of the Specification dated June 1999 relating to “Concretor, including Reinforcing and Formwork”

The National Institute of Water and Atmospheric Research Report on the depth and density of snow on the ground surrounding the Stadium a day after the collapse (“the NIWA Report”) (Hendrikx 2010).

DETERMINATION OF COLLAPSE SEQUENCE

To determine the cause of the collapse required careful observation of the way the collapse debris lay after the collapse, laboratory examination and mechanical testing of materials and components salvaged from the site, identification of the collapse sequence, computer based 3D structural analysis of the structure under snow loadings and structural calculations.

Demand-to-capacity ratios for critical components were calculated in three steps. The first using specified material and section properties and configurations. The second used as-built section properties and configurations. The third used as-built section properties and configurations, with materials properties derived from the laboratory testing.

The critical components were then ranked in terms of demand to failure capacity ratio for a uniform snow loading at which the most critical component was calculated to have developed a demand to capacity ratio of 1.0 initiating collapse.


REPORT METHODOLOGY continued


SITE EXAMINATION

The authors visited the Stadium site following the collapse on 24 and 25 September 2010 and examined the debris remaining from the collapse at that time (“the Site Examination”). Photos taken by others prior to debris being moved and prior to the Site Examination have also been considered. The authors were advised that the condition and orientation of the debris remaining on site at the time of the Site Examination was the same as it had been immediately after the collapse.

LABORATORY EXAMINATION

Portions of salvaged roof trusses at critical failure locations were selected during the Site Examination by the authors.

Detailed examination of welds in these items was performed by Materials and Testing Laboratories Ltd (“MTL”). Examination of fracture surfaces, and mechanical testing was performed by HCL in conjunction with Uniservices Ltd (“Uniservices”) at the University of Auckland (“the UoA”).

The results of the laboratory examination and mechanical testing are summarised in Appendix C of this report and reported in detail in a separate HCL report (Hyland 2010) (“the Laboratory Examination”).



3 STADIUM DESCRIPTION

STADIUM LOCATION

Stadium Southland is located in Surrey Park, on Surrey Park Road, Invercargill.

STADIUM FACILITIES

The Stadium provided a 9108 m2 indoor sport facility for the Southland region. It housed five multi-purpose Community Courts, two Events Courts, four squash courts, a climbing wall, seating for 2566 people, full service amenities, a lounge, bar and kitchen, and parking for 700 cars (Figure 4).

The project commenced construction in 1999 and was opened in 2000.


STADIUM DESCRIPTION continued


STRUCTURAL DESCRIPTION

The following is a description of the Stadium roof structure as it stood just prior to the collapse on 18 September 2010. This description is based on the Drawings, observations on site after the collapse, subsequent examination of structural remnants, and photographs by others.

Only those parts of the roof structure relevant to the collapse on 18 September 2010 are described. A three-dimensional view of the main roof structure is shown in Figure 51 (in Appendix D). The various components have been labelled.

In general the roof structure clear spanned over two large inter-linked spaces, namely the ‘Community Courts’ and the ‘Events Courts’. The Community Courts space measured 96.35m x 36.6m and the Events Courts space measured 46.1m x 39m on plan.

Throughout this entire area the roof comprised 0.55mm thick profiled metal sheet cladding, on safety netting and building paper, supported on proprietary cold formed steel purlins with two lines of purlin braces in each bay. Beneath the purlins were top-hat section strong backs supporting insulation and a screw fixed 16mm thick medium density fibreboard (MDF) ceiling. The purlins spanned in the east-west direction above the Community Courts and in the north-south direction above the Events Courts.

Above the Community Courts, along each of the roof truss lines there was a 2m wide continuous glazed skylight. The roofing and ceiling stopped at each side of these skylights with purlins continuing through beneath the glass to connect to the supporting trusses at mid-height as shown in Figure 5.


STADIUM DESCRIPTION continued


The purlins were supported by steel roof trusses fabricated from square hollow section (SHS) and rectangular hollow section (RHS) members.

The Community Courts roof trusses spanned 36.6m between Grid 4 and Grid 5. These trusses are referred to as T1 to T9 starting from the eastern end.

Along Grid 4, the Community Courts roof trusses T1 to T5 were supported by truss T10 of the spine truss, which in turn spanned between concrete columns C16 and C14 on Grids E and F respectively. Further along Grid 4, between Grids E and A, the remaining Community Courts trusses T6 to T9 were supported by a framework of steel beams and posts restrained at mid-height by the edge of the concrete mezzanine ‘upper’ floor.

Along Grid 5 on the south wall the Community Courts roof trusses were supported on concrete columns that were connected into concrete wall panels. The walls at each end of the Community Courts, ie at Grids A and G comprised part-height concrete wall panels with lightweight structural steel framing and cladding above. These end walls each supported one end of the roof purlins.

Above the Events Courts the purlins spanned north-south and were supported by secondary trusses numbered T16 to T24 that each spanned 23m in the east-west direction. Supporting these secondary trusses at the eastern end is the wall on Grid F. This wall was part height concrete with lightweight steel framing and cladding above.

Above the centre of the Events Courts, and at the western end, the secondary roof trusses were supported by primary trusses T12 to T15 that spanned 42m in the north-south direction. These trusses were supported by concrete columns C17 to C20 at the northern side of the building and by spine truss T11 at Grid 3.

‘Lean-to portal frames’ T1 to T6 spanned east-west between Grids B and C, adding further load into the Events Courts primary truss T15.

Spine trusses T10 and T11 along Grids 3.5 and 4 between E and F, and supporting concrete columns C13 to C16 were key elements in the roof structure. They supported one end of trusses T1 to T5 over the Community Courts and also one end of trusses T12 to T15 over the Events Courts. The tributary area supported by the spine trusses was approximately half the entire Events Courts roof and one quarter of the entire Community Courts roof.



4 REMEDIAL WORK UNDERTAKEN DURING CONSTRUCTION

Some changes were made to the structural design after the steel fabrication tender had been accepted to meet budget constraints. These included reducing the steel tube wall thickness of the top and bottom chords of the Community Courts trusses T1 to T5, from 9 mm to 6 mm thick.

Council records showed that excessive deflections were observed in the roof trusses above the Community Courts area and cracking of the supporting concrete column heads at the south wall during construction of the Stadium.

The resulting Structural Review Report of December 1999 (“the 1999 Review Report”) identified the following structural members and connections that did not meet design capacities required by the relevant Standards at the time:

Community Courts roof trusses

Community Courts truss to column connections on Grid 5

Trusses T10 to T15 connections to the concrete columns C13 to C20

In addition concerns were expressed in the 1999 Review Report about the slenderness of the concrete columns C13 to C16.

Remedial strengthening works required as a result of the 1999 Review Report involved strengthening and pre-setting the Community Courts roof trusses, the modified shoe details supporting roof trusses T12 to T15 onto concrete columns, epoxy grouting cracks in columns C13 to C16 and pre-cast panel top connections along Grids A and G between Grids 4 and 5. These remedial works were shown on Eng Drawing 97139 Revision A dated December 1999 (Figure 52 in Appendix D).

In the original design the connections between the Community Courts roof trusses and the tops of the Grid 5 columns had been designed as ‘rigid’ to minimise deflections and mid-span bending moments in the trusses. However, these connections, and the tops of the concrete columns, were found during construction to be unable to withstand the resulting rigid joint actions. As a result these connections were required to be modified by removing the connecting bolts on the column faces and providing a gap between the column faces and the truss bottom chord. This provided a nominally pinned support at the end of the trusses to the Grid 5 columns.

The loss of bending resistance from the Grid 5 truss-to-column connections was to be compensated for by strengthening the chords over the middle half of the trusses. However this also meant that failure only needed to


REMEDIAL WORK UNDERTAKEN DURING CONSTRUCTION continued


occur at the mid-span of a truss for a collapse of a truss to occur, rather than at both mid-span and at the Grid 5 column head. Even so collapse of a single truss would not normally be expected to result in the progressive collapse of the roof as a whole, as occurred in the Stadium.

The specified remedial procedure to re-camber and strengthen the Community Courts roof trusses, as outlined shown in Figure 52(in Appendix D) included the following steps:

1. Jack trusses to achieve approximately flat soffit.

2. Remove all bolts connecting trusses to columns on Grid 5.

3. Cut cleats securing bottom chord of trusses to supporting truss on Grid 4.

4. Cut top chords at quarter points and mid-span and continue jacking to achieve a 230mm upwards camber.

5. Prepare cut edges of SHS (top chord) members for complete penetration butt welds and insert mild steel plates of appropriate thickness. Weld all-around and grind flush on both vertical faces of the chord at the mid-span splice only.

6. On trusses with 125x125x6 mm SHS chords, plate top and bottom chords, both sides, with 100x10 mm Grade 300 mild steel flats. Fit the plates to the chord profile by complete penetration butt welding segments end to end.

7. Also add strengthening plates to four specific web members per truss.

8. Place steel shims between the top of the Grid 5 support columns and the top chords of the Community Courts trusses, no closer than 250mm from the inside face of the columns.

9. De-prop the trusses and make good the fixings at the top of the columns and at Grid 4.

10. Repair existing cracks to concrete columns using epoxy grout injection and install proprietary packing.

In the opinion of the authors this work was complex and would have been difficult to execute to the requirements of the Steel Structures Standard. It would have required the removal of the glazing panels above the Community Courts trusses.



5 SNOWFALL DESCRIPTION

Photos taken by members of the public on the Saturday 18 September 2010 indicate snow thicknesses of approximately 75 to 150 mm (Figure 7).

Extracts from the NIWA Report (Hendrikx 2010) relating to the Stadium roof collapse include:

“The snow event in Invercargill and across Southland started on the evening of 17 September and included 16 hours of heavy snowfall, followed by intermittent snow, rain, hail and sleet, prior to the measurements being taken on 19 September.”

“The majority of the snowfall is thought to have occurred from approximately 8 pm on 17 September to around 12 noon on 18 September (the approximate time that the stadium roof collapsed), with the most intense snowfall occurring after 6 am on 18 September. Further snow showers did occur after this time period in the hours and days after the main event, but the majority of snowfall occurred in this initial 16 hour period. This is our estimate, based on the available meteorological data, media reports and discussions with the local meteorological observer…”

NIWA used credible scientific techniques to measure the depth and density of the snow on the ground adjacent to the stadium and therefore provides the best measure available of the snow on ground load.


SNOWFALL DESCRIPTION continued


The NIWA Report stated that the snow adjacent to the stadium on 19 September, (ie the day after the collapse), had a mean depth of 120 mm and mean density of 3.7 kN/m3, representing a snow on ground load (Sg) of 0.45 kPa (or kN/m2). The snowfall was intermixed with hail. The density measured was higher than the density of 2.9 kN/m3 assumed for sub-alpine snow in the snow loadings standard AS/NZS 1170.3:2003.

Varying reports of snow levels immediately prior to the collapse were made by members of the public. The snow storm was considered by some to be the worst in living memory.

The snow on ground load for a 1 in 250 year design event for Invercargill in the Snow Loadings Standard AS/NZS 1170.3:2003 was 0.51 kPa which is a little higher than the 0.45 kPa measured the day after the collapse.

The NIWA Report stated that there was little or no wind at the time of the collapse at noon on Saturday 18 September, 2010. This average wind speed measured was 3 m/s with peak 3 second gust wind gust of 6 m/s (Figure 6).

The wind speeds measured were typically below the threshold for wet snow to be transported by the wind. As a consequence the NIWA report states that “…substantial blowing snow and redistribution of snow seems to have been unlikely in this event.” Uniform loading therefore appears to have been the most likely scenario rather than drifts occurring against the south side of the spine truss up-stand on the roof.


SNOWFALL DESCRIPTION continued


It is not possible to make an accurate assessment of the snow load on the Stadium roof was when it collapsed solely on the basis of the NIWA Report. There are two reasons for this.

The first reason is that the snow depth and density was measured one full day after the roof collapsed. Intermittent snow, rain, hail and sleet continued to fall during that time that may have increased the density of the snow on ground. The collapse of a Shop Roof close to the Invercargill city centre one day after the Stadium roof collapse, is evidence that the snow load had continued to increase up until the time the measurements were made. The wind speeds at the time of the Shop Roof collapse were not much greater than those at the time of the collapse of the Stadium roof.

The second reason is that it is recognised in the loadings standards that snow on ground depth and densities and those on roofs vary by a factor related to the slope of the roof and wind speed at the time. There is some doubt about the accuracy of those relationships for sub-alpine locations in New Zealand like Invercargill. The relative warmth of the roof due to warmer temperatures inside the Stadium compared to the ground may also have reduced the snow loading.

The commentary to the Snow Loadings Standard AS/NZS 1170.3 Supp 1:2003 notes that “usually there is significantly less snow on a roof than on the ground”. It also states that “The basic relationship between roof snow load and ground snow load has been set at 0.7 for a sheltered flat, or nearly flat, roof. This is lower than the values of 0.8 and 0.84 given in the European, Canadian and American Standards. The lower value reflects the different conditions in Australia and New Zealand.” The average daily temperatures in New Zealand during the winter are relatively high due to the moderating influence of the surrounding oceans compared to northern hemisphere nations with large continental land masses. The authors support a review of this relationship being undertaken, as recommended by the NIWA report.

On balance, the authors consider that the actual equivalent uniform snow load on the roof at the time when it collapsed would have been less than 0.45 kPa, based on the NIWA Report, and is estimated to have been 0.30 kPa based on the structural analyses undertaken for this report using the critical truss chord splice capacity determined by testing.



6 COLLAPSE DESCRIPTION

The roof on Stadium Southland collapsed at around noon on 18 September 2010 following a heavy snowfall.

An aerial photograph of the Stadium was published in the Southland Times the day after the roof had collapsed (Figure 2). The Community Courts are at the right hand side in the photograph and the Events Courts are at top left. The direction north is towards the left.

The areas of the Stadium that collapsed included a portion of the end walls, roof purlins and roof trusses above the Community Courts; some roof purlins and one end of the main roof trusses above the Events Courts; and the main roof ‘spine’ trusses and supporting columns that were located between the Community Courts and the Events Courts. Concrete columns C13 and C14 remained standing immediately after the collapse but were demolished for safety reasons along with the attached access stairway soon after (Figure 8).

It was reported that several tennis players who had been using the Community Courts that morning had just left the building when the roof collapsed. They described hearing a loud ‘crack’ or explosive sound and seeing the doors at the eastern end of the Community Courts blow open with the pressure and a roof panel fly past nearby. The portion of the roof at the western end of the Community Courts then collapsed. Trusses T6 to T8 and a portion of the roof they supported remained in place.

Access Tower and Spine Truss support columns C13 and C14


COLLAPSE DESCRIPTION continued


Although there was one person in the building at the time of the collapse he was not in the area affected by the collapse. There were no injuries or loss of life as a result of the collapse.



7 CONTEXT OF THE STADIUM SOUTHLAND COLLAPSE

BACKGROUND

There were a number of other structures in and around Invercargill that collapsed or suffered damage from the snowfall. Two of these buildings were visited by the authors on 25 September 2010.

SHOP ROOF COLLAPSE

A shop in the commercial area of Invercargill suffered a total collapse of its recently built light gauged steel tube truss roof (“the Shop Roof”) the day after the Stadium roof collapsed (Figure 9). This is evidence that snow loads had continued to increase following the Stadium roof collapse the day before. The wind speeds measured at the time of both collapses were similar (Figure 6).

The pitched roof trusses were bolted by end plates to nominally pinned-base steel universal beam section columns at one end and reinforced concrete pilasters at the other end. The trusses had collapsed by formation of hinges from top chord compression buckling in the mid-span panels, and variations of bottom chord buckling or top chord tension failure adjacent to the column faces at each end, combined with steel column rotation about their bases. Some cracking had also developed in the concrete walls where the frames had pulled in from the walls as the roof collapsed.


CONTEXT OF THE STADIUM SOUTHLAND COLLAPSE continued


FLY BRACE RESTRAINT BUCKLING IN LIGHT INDUSTRIAL BUILDING

A further indication of the level of loading experienced from the snow fall was seen in a light industrial steel portal frame building where permanent buckling deformation had occurred to tension-only fly braces. These were required to provide lateral restraint to the bottom chord of the roof rafters and the supporting columns (Figure 10). The purlins showed no obvious signs of permanent set.

There appeared to have been some lateral displacement of the bottom flanges of some rafter elements at column support locations. This is consistent with a reduction of lateral buckling restraint to the bottom flanges of the rafters at the supports as the fly-braces themselves buckled as they resisted the snow induced downwards displacement of the purlins each side of the rafters.



8 STADIUM SOUTHLAND SITE EXAMINATION

The aerial photo (Figure 2) on Sunday 19 September 2010 shows how the roof structure lay soon after the collapse. However the access stairway to the spine trusses and columns C13 and C14 had been deconstructed by that time for safety reasons (Figure 8).

DBH staff visited the stadium site on Tuesday 21 September 2010. At that stage very little debris had been removed from the site. Photos taken during that visit were made available for this report (Figure 14 in Appendix B).

The authors of this report conducted a Site Examination on 24 and 25 September 2010 (“the Site Examination”). This commenced late on Friday 24 September 2010. Trusses T1 to T3 and associated roof debris had been removed and stacked outside, and trusses T1 and T2 had been taken to the off-site debris storage yard by that stage. The authors were advised by the engineer acting for the insurer (“the Insurer’s Engineer”) at the time of the Site Examination that the condition of the remaining debris and the way it lay was the same as it had immediately after the collapse.

On Saturday 25 September the damaged heads of the south wall Grid 5 columns C10 and C11 that had supported trusses T1 and T2 were inspected using a cherry picker (Figure 11). The remainder of the collapsed structure was inspected at ground level. Access was not permitted below trusses T6, T7 and T8 that had not collapsed. However, access was able to be gained at the western end of the Community Courts area to examine the collapsed truss T9.

Trusses T1 and T2 had been removed from site and laid out on the ground along with portions of demolished concrete columns C13 and C14 in the off-site storage yard. These were also examined late on 25 September, 2010.



9 LABORATORY EXAMINATION

Portions of the salvaged trusses T1, T4 and T9 at the locations of fracture and failure were cut out for detailed examination of welds and fracture surfaces, and tensile testing in Auckland. Samples of threaded rod that connected trusses T1 to T5 onto truss T10 were included as well as a portion of concrete column C15 that had supported truss T11.

The column remnants were used to determine the compressive strength of the concrete and tensile strength of the reinforcing steel and TCM assemblies.

The Laboratory Examination was conducted by HCL, in conjunction with Uniservices and MTL, and commenced on 26 October 2010. This was the date of delivery of the crate of salvaged items from Invercargill to the University of Auckland Engineering Workshop and Laboratories (Figure 12).

A detailed report on the Laboratory Examination was prepared by HCL in December 2010 (Hyland 2010). A copy of that report was disclosed to the Insurer who had assisted with extracting samples for the Laboratory Examination and with its cost. A summary of the results of the Laboratory Examination is included in Appendix C.

A range of procedures were employed to determine mechanical properties of materials, as-built assembly capacities, assess compliance of the components with the relevant standards, design documentation or accepted good practice. The range of examination and testing methods included: welding inspection, hardness testing, tensile testing of steel, chord strut assembly compression testing; and concrete core testing.

Compression tests of fabricated truss top chord assemblies were conducted to determine their collapse mode and capacity as affected by the observed 3-sided end bearing at the splice locations that was observed on the Community Courts trusses.


LABORATORY EXAMINATION continued




10 EVALUATION OF THE STRUCTURE PRIOR TO COLLAPSE

INTRODUCTION

The purpose of the structural condition evaluation by the authors was to:

1. Determine the effect of the remedial strengthening work undertaken during the original construction as-drawn and as-built, on the structural capacity of critical components.

2. Identify the component that initiated the collapse by determining which component had the highest demand to failure capacity ratio under a uniformly applied snow loading.

METHODOLOGY

The basis of the 3D computer modelling, structural analysis and wind loadings used is described in Appendix A. A comprehensive selection of photographs of the critical structural components on site is also provided within Appendix B. Further detailed photographs of failed components taken during the Laboratory Examination are provided in Appendix C and the HCL Laboratory Examination report.

The evaluation of the structure included a Design Basis Review in which components were assessed based firstly on the Drawings, including the remedial work that had been required following the 1999 Review Report. This review used specified section and material properties, and strength reduction factors to determine design capacities. A design demand to capacity ratio greater than 1.0 As-Drawn indicated design non-conformance with the relevant standards.

Secondly the as-built configuration was evaluated for fitness for purpose using the measured section sizes, measured weld design throat thicknesses (“DTT”) and actual connection configurations, but also using specified material properties and strength reduction factors used in design. The specified material properties are typically the lower 5 percentile strength expected for that class of material. An As-Built design demand to capacity ratio greater than 1.0 indicated As-Built design non-conformance with the relevant standards.

Neither the As-Drawn or As-Built design demand to capacity ratios calculated for the Design Basis Review (Table 1) represented the actual situation at collapse.

A Failure Basis Review was then done. This review comprised a detailed review of site observations and photos in conjunction with calculations to assess collapse process, actions and capacities for each component. The calculations for this failure analysis based review used a uniform snow load of 0.30 kPa on the roof, without load factors, and realistic expected values


EVALUATION OF THE STRUCTURE PRIOR TO COLLAPSE continued


for material properties based on test results from the Laboratory Examination, to calculate failure demand to capacity ratios for the components at the time of the collapse. A failure demand to capacity ratio of 1.0 indicated failure initiation in this case.

A 0.30 kPa uniform snow load was the load at which the failure demand to capacity ratio of the critical truss T1 top chord exceeded 1.0, using the chord splice capacity determined from testing. This included an assessment of the actual self-weight of the roof, steelwork, ceiling and fittings at the time of the collapse. At this demand to capacity ratio collapse was calculated to have initiated in the T1 top chord at midspan. The buckling capacity of the top chord was found by analysis to be sufficient to prevent a buckling induced failure to occur in the top chord prior to splice failure.

A comparison of the failure demand to capacity ratios shown in Table 1 enabled the identification of the most critical components in the collapse. It also highlighted the often unrecognised safety margins provided for in the materials supply standards for bolts and steel that in this case prevented collapse occurring at lower snow loads.

APPLICABLE STRUCTURAL DESIGN STANDARDS

The original design calculations were not made available for this report. The loadings and materials standards adopted and used in the design of the Stadium are therefore not certain. However the date of the design suggests that it was likely to have been carried out in accordance with standards which were applicable in 1999. These have been used in the structural analyses undertaken for this report. Those Standards (“the Standards”) were as follows:

NZS 4203:1992 – The Loadings Standard

NZS 3404:1997 – The Steel Structures Standard;

AS/NZS1554.1:1995 –Steel Structures Welding Standard;

NZS 3101:1995 – The Concrete Structures Standard

Reference has also been made to AS/NZS 1170.3:2003 Snow Loadings and the NZBC Amendment 9 of December 2008.




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SUMMARY OF FINDINGS FROM THE STRUCTURAL EVALUATION

Community Courts Roof Trusses T1 to T9

The Community Courts roof trusses T1 to T9 had originally been designed to rely on end-fixity at the tops of the Grid 5 columns along the south wall. All of these roof trusses were modified on-site following the 1999 Review Report. Side plates were welded to the truss chords with the intention that the as-drawn design capacity ratios then became acceptable under the Standards applicable in 1999. However, the Site Examination and the Laboratory Examination found that the fabrication fit-up and welding of the top chord compression splices and the side plates did not conform with the Drawings or with the applicable Standards (Figure 13).

The side plates were not continuous over the top chord splices. The splice plates were not welded end to end with complete penetration butt welds. The top chord splices were typically only welded on three sides indicating that access to the top face of the top chord for welding was restricted. The splice packer plates were missing from some of the splice locations in the top chords at quarter points. Packer plates at the top chord splices at midspan locations in the top chord had only three sided end bearing.

The connection of truss T1 to the spine truss T10 was also compromised by the site welding of the bottom chord cleat that prevented movement of the truss T1 bottom chord. This increased the demands on the threaded rod in the top chord connection.

South Wall Grid 5 Column Heads

The south wall column heads were modified as part of the remedial strengthening works to provide simple support conditions for the Community Courts trusses. The Site Examination found the heads of columns C3, and C7 to C11 had sheared off during the collapse of the Community Courts trusses. Some bolts appeared to be missing from the connections of the trusses to the tops of the concrete columns on the south wall.

Connection of the Spine Trusses to the Concrete Columns

The 1999 Review Report identified the need for “modified shoe details” for trusses connected to columns C13 to C16 and for a minimum of four effective TCM connections to be provided into columns C13, C14, C15 and C16.

During the Site Examination it was found that the ends of Trusses T10 and T11 had only one packing plate welded in position and other mild steel packing plates were un-welded. This reduced the capacity of the connection by increasing secondary bending actions in the bolts. The as-built arrangements for the fixings were found by calculation to have a design to demand capacity ratio exceeding 1.0, meaning that they were non-complying after construction.




Spine Truss Support Columns C13 to C16

The spine trusses support columns were slender. The axial load carrying capacity of column C15 was found to be sensitive to the level of lateral restraint provided by the Control Room floor. As can be seen from Items 1 and 2 in Table 1 restraint by the Control Room floor may have improved the buckling design demand to capacity ratio of column C15 sufficiently so that it could conform with design requirements.

The Concrete Structures Standard NZS 3101:1995 gave limited guidance on the assessment of effective lengths for columns as slender as these.

A buckling analysis using the widely accepted G-Factor method (Bridge and Fraser 1987) found the columns to be adequately restrained to prevent weak axis buckling in the north-south direction for the design basis review if the Control Room floor was considered to provide lateral restraint. The

54

Changes to all Community Courts trusses T1 to T9

Cut top chord

Install continuous side plates to chords

Add plates to diagonal ties at each end

Remove end connections

Reinstall bolts but not to column face Splice top chord

Reinstate end connections

Jack-up trusses

Pre-set trusses

Defects in Community Courts trusses

54

Some bolts missing

3- sided fit up and welding, no backing bar sleeves, no packers in some, undersized welds

No butt weld connecting ends of strengthening plates No backing bar sleeves, bevelled steel

fit-up plugs, undersized welds

Strengthening plates not continuous over splice

Some chord and web tube thickness less than specified

Movement prevented in Truss T1 cleat by welding




level of restraint action required was 2.5% of the axial compression action in the column, which was in the order of 17 kN.

It is not obvious from the Drawings whether it had been intended for the Control Room floor to provide this level of lateral restraint. The floor was shown on the Drawings as screwed and glued to floor joists which were bolted to steel 230 PFC sections that were bolted to the columns and adjacent walls each end with 2 –M16 anchors. In shear these would be expected to have a total design capacity of at least 58 kN each, which was greater than the required lateral restraining action to the column. So some level of restraint could be expected from the Control Room floor.

For the failure basis evaluation it was found that restraint by the Control Room was not necessary to prevent weak axis buckling of C15 (Table 1).

The shear and confining reinforcement detailing of columns C13 to C16 was found not to conform with the minimum spacing provisions in the concrete Standard, though the design demand to capacity ratios for shear on the columns was calculated to be less than 1.0. This indicated that the columns had adequate shear capacity even though the minimum shear reinforcing requirements had not been met.

The 1999 Structural Review commented on the eccentricity of the spine truss connections to columns, the slenderness of columns and cracking that was observed in the columns at that time. The Laboratory Examination of the C15 column remnant found that cracks in the column had been filled with epoxy grout (Figure 45 in Appendix C).

Roof Purlins

The Drawings had conflicting information regarding purlin thickness which raised questions about whether the longer span end-bay purlins (between truss T1 and Grid G and between truss T9 and the western wall) had been up-graded to a thickness of DHS 250/18 or not. However the authors were advised that measurements of remnants of the purlins by the Insurer’s Engineer confirmed that the end bay purlins had been the heavier gauge DHS 250/18.



11 CONSTRUCTION MONITORING, WELDING AND INSPECTION REQUIREMENTS

The requirements of the relevant Standards applicable at the time for construction monitoring, welding and inspection relevant to the Stadium roof construction were as follows:

1. NZS 3404:1997 Section 1.6.3 set out requirements for construction review (or construction monitoring as it is referred to in IPENZ / ACENZ guidelines).

Clause 1.6.3.1 reads:

“All stages of construction of a structure or the part of a structure to which this Standard is applied shall be adequately reviewed by a person who, on the basis of experience or qualification, is competent to undertake the review. (This person is termed the construction reviewer).”

And 1.6.3.2:

“The extent of review to be undertaken shall be nominated by the design engineer, taking into account those materials and workmanship factors which are likely to influence the ability of the finished construction to perform in the predicted manner.”

2. There are four notes following the above clauses in the Standard as follows:

A construction reviewer might be a registered engineer with suitable experience, or a building certifier.

Welding supervisors shall be qualified in accordance with clause 4.11.1 of AS/NZS1554.1 or clause 4.11.1 of AS/NZS 1554.5, whichever is appropriate.

Painting inspectors should hold Certified Coatings Inspector qualification.

Clause 4.1 of either AS/NZS 1554.1 or AS/NZS 1554.5 requires establishment of welding procedure sheets by the fabricator. Prior to the commencement of welding, these shall be approved for the particular job by the design engineer, or by the construction reviewer, or by his/her nominated representative.


CONSTRUCTION MONITORING, WELDING AND INSPECTION REQUIREMENTS continued


3. Inspection of welding was covered in Appendix D of NZS 3404:1997. Clause D2 covers the extent of non-destructive Examination for welds and reads:

“The extent of non-destructive examination required shall be determined by the Principal to the approval of the Engineer.”

Further guidance was then given in Table D1 of NZS 3404: 1997, however it is noted that Table D1 formed an informative part of the Standard and so was not mandatory.

Clause D2 has been identified as a weakness in NZS 3404:1997 because the option could have been taken by the Principal not to carry out any non-destructive examination of welds at all.

The current edition of NZS 3404.1:2009 has modified these provisions to require a minimum extent of non-destructive examination (NDE) of welds as a function of weld failure consequence and weld demand level.

Under this current edition of the Standard the minimum extent of non-destructive examination would have been 100% visual scanning, between 25% and 100% visual examination and between 0 and 10% other NDE (ultrasonic, magnetic particle, liquid penetrant).



12 COLLAPSE SEQUENCE

The authors have concluded that the collapse initiated with compression peeling and crushing failure at the mid-span top chord site splice in truss T1 at the eastern end of the Community Courts. Similar top chord compression failures consequently developed in trusses T2 to T5, causing a westward displacement of the roof to occur (Figure 53 in Appendix D).

This pulled the connected spine trusses T10 and T11 and their supporting columns westward leading in the development of flexural hinging in concrete columns C15 and C16. These hinges developed immediately below trusses T10 and T11, and at the Control Room floor level.

As the westward displacements increased the bottom chord bolts connecting trusses T10 and T11 to columns C13 and C14 fractured in combined tension and shear. The roof then fell to the ground as columns C14 and C15 collapsed and the remaining bolts holding trusses T10 and T11 to columns C13 and C14 sheared off. It is likely that the Community Courts trusses T1 to T5 were simultaneously pulled off their column head supports at the south wall. These column head connections were fragile and lacked development of the connectors and reinforcement to prevent them shearing the column heads.

During compression testing of mock ups of the top chords of the trusses, the peeling and crushing failure of the chord occurred without significant noise. The tensile fracture of steel bolts and rods during testing however was accompanied by a loud explosive noise. The collapse is also expected to have occurred rapidly once it initiated. Fracture of Community Courts trusses bottom chord welds, pullout of roof bracing rods from truss vertical struts, fracture of spine trusses T10 and T11bottom chord bolts at columns C13 and C14, and concrete column fractures of C15 and C16 are likely to have occurred almost simultaneously and together could explain the loud exploding noise reported.

Truss T9 then fell to the floor as the bottom chord weld fractured and the strengthening plates yielded, and then broke the welds attaching it to the chord. This appears to have been initiated by the collapse of the Events Courts and Community Courts roof east of Truss T6 which caused a sudden build up of air pressure in the west end of the Community Courts. This pressure was reported by those who had just left the building to have been sufficient to blow out the end walls and doors at the east end of the Stadium. This may have forced the roof upwards sufficiently to fracture the top chord mid-span site splice of truss T9, which had less than 1 mm DTT weld thickness connecting the chord to the splice (Figure 35), prior to the bottom chord fracture occurring.

The failure capacity ratios in Table 1 support the conclusion that the collapse of the roof structure was initiated by compression failure of the mid-span top chord in roof truss T1 at the eastern end of the Community Courts.


COLLAPSE SEQUENCE continued


The purlins reportedly installed in the end bays and those installed elsewhere were found to have adequate capacity to sustain the snow loads at which collapse is estimated to have occurred.

Community court trusses T6, T7, T8 and T9 were not supported by spine truss T10 and had thicker 9 mm wall thicknesses in their top and bottom chords. While it is likely that T6, T7 and T8 all had similar weaknesses as T9, they were evidently able to resist the pull down effect transmitted through the purlins from the collapsing roof both sides and remained in place.

It was noted that trusses T1 and T9 supported a greater tributary width of roof than the other Community Courts trusses. When considered together with the varying steel tube wall thicknesses of the truss chords and fabrication deficiencies, this explains why both ends of the roof collapsed rather than the roof supported by trusses T6, T7 and T8.



13 REASONS FOR THE COLLAPSE

The authors consider from their study the reasons for the collapse to be as follows:

1. Defects in the Community Courts trusses, their connections, and those of the spine trusses T10 and T11 to their supporting columns C13 and C14, caused significant weakening of the structure. The roof structure otherwise should have been able to support the snow loads that occurred on the weekend of 18 September 2010.

2. The following specific defects in critical components contributed to the collapse :

The top chords of truss T1 to T5 at the mid-span splice locations had insufficient end bearing and welding at the splice plates.

The quarter point splices of trusses T4 and T5 did not have end bearing splice plates installed and had insufficient welding.

No welding occurred to the top surface of the top chords of the Community Courts trusses at most splice locations.

The top chord side strengthening plates were not installed continuously past the chord splice locations and were not joined by complete penetration welds. Welds attaching the plates were applied to painted surfaces.

The bolted connection of spine truss T10 to column C14 and T11 to C13 had unwelded packing plates increasing the stresses on the bolts from secondary effects.

The connections of the Community Courts trusses to the south wall concrete columns were brittle and were unable to prevent the trusses being pulled off their supports.

The mid-span bottom chord site welded splice in all the trusses examined had been incorrectly prepared and welded. Complete penetration welds were not achieved as specified.


REASONS FOR THE COLLAPSE continued


3. The average ground snow load measured by NIWA one day after the Stadium collapse was 0.45 kPa. Snow on ground loads are considered by the loadings standard AS/NZS 1170.3:2003 to be higher than those that occur on sloping roofs. It is also seems that the snow load may have increased from the time when the Stadium roof collapsed to when the NIWA measurements were taken. Sleet, rain and some snow continued to fall in the day following the Stadium roof collapse. The collapse of a recently built steel truss Shop Roof on another site, one day after the collapse of the Stadium roof in similar weather conditions and winds, is evidence of this increase of snow load.

4. The snow load on the roof at the time that it collapsed was estimated by the authors to have been as low as 0.30 kPa. This was based on top chord compression capacities found by testing during the Laboratory Examination. The requirements for the building when it was originally designed were for a factored roof snow load of 0.40 kPa. Therefore the snow load alone does not explain the collapse.

5. The structure was susceptible to progressive collapse due to a number of factors including:

Use of thinner truss chord tube wall thickness for Community Courts trusses T1 to T5. This seems to have caused the collapsing trusses T1 to T5 to be dragged westwards towards the stronger trusses T6 to T8 with the thicker tube wall thickness which remained standing.

Numerous weakening defects in similar locations in trusses T1 to T5, and T9.

Design modifications that resulted in simply supported trusses over the Community Courts.

Site splices at the most highly stressed mid-span locations of the Community Courts trusses.

Inability of the column head connections at Grid 5 on the south wall to prevent the Community Courts trusses pulling off as a failure developed in trusses T1 to T5.

The spine truss support structure was laterally flexible and did not have the stiffness, strength or bracing back to sufficiently rigid and strong restraints, to resist the


REASONS FOR THE COLLAPSE continued


westward drift imposed on it as the Community Courts trusses began to collapse.

6. The spacing of steel reinforcing ties in the concrete columns C13, C14, C15 and C16 supporting the roof spine trusses T10 and T11 were less than required by the applicable Concrete Structures Standard NZS 3101:1995.

7. At a late stage of construction Council records show that modifications were designed and carried out to resolve problems with excessive deflections and the design capacity of the Community Courts trusses following an earlier design change, south wall column head connections, spine truss connections to columns and cracking of spine truss columns.

Portions of the on-site steel fabrication and welding for the remedial works were not compliant, in ways material to the collapse, with the Drawings, and/or the Steel Structures Standard and/or the Welding Standard.

The remedial strengthening works specified were complex and would have been difficult to achieve.

8. Threaded rods connecting the Community Courts trusses to spine truss T10 and the bolts connecting spine trusses T10 and T11 to column C13 to C16 were found in the Laboratory Examination to be significantly stronger than was specified. This may have prevented collapse initiating at these locations at lower snow loads.



14 RECOMMENDATIONS

SUMMARY RECOMMENDATIONS

The authors make the following recommendations to prevent similar collapses happening to other public buildings in areas subject to snow loadings. A number of these would also be applicable to buildings in other locations not subject to snow:

Design of Buildings Subject to Snow Loading

1. Research be undertaken to improve understanding of snow loadings for New Zealand coastal sub-alpine areas.

2. The relationship between roof and ground snow loads due to the effect of roof slope in New Zealand sub-alpine areas be confirmed by research.

3. Review the statistical reliability basis of snow loadings, return periods and corresponding load and materials strength safety factors for the loadings, and the steel and concrete structures standards.

4. Review the adequacy of the extent of concrete column buckling provisions in the Concrete Structures Standard NZS 3101.

5. Develop guidelines for the design of roof structures subject to snow loadings to prevent progressive collapse in the event of snow overload. Require the connections of roof rafters and trusses to supports, to be sufficient to prevent them being pulled from their supports should those members suffer overload failure.

6. Require that the columns and walls supporting roof structures subject to snow loading be sufficiently stiff, strong or braced to resist transverse drifts imposed by collapsing rafters under overload conditions, thereby preventing progressive collapse.

7. Require structural engineers to identify on the drawings collapse critical components of important roof structures subject to snow loading, for particular attention by reviewers, constructors and construction monitors.

8. Mandatory levels be set for independent third party design review by appropriately qualified people for steel roof structures of public buildings subject to snow loadings.


RECOMMENDATIONS continued


Construction of Buildings Subject to Snow Loading

1. Mandatory minimum levels of competency be set for companies and key personnel undertaking the construction of steel roof structures of public buildings subject to snow loadings.

2. Mandatory levels be set for independent third party construction monitoring by appropriately qualified people for steel roof structures of public buildings subject to snow loadings.

3. Ensure that public buildings expected to be affected by snow loadings are not opened to the public for use until the adequacy of the construction of all roof structural components have been appropriately inspected and certified.

4. Recommend that snow overload alarms be installed in public buildings with large span roofs subject to snow to warn occupants to exit the buildings when snow loads exceed the specified design limits.

DETAILED RECOMMENDATIONS

Joint Australian and New Zealand Loadings Standard AS/NZS1170

General Structural Design Provisions

1. Require the consideration by design engineers of a collapse limit state in the design of important structures under snow actions. This would include consideration of the potential collapse mechanism and the relative level of redundancy and ductility in the system. A structural performance factor applied to loadings used could be introduced similar to that used in seismic design considerations.

2. Structural systems with high resilience that require the development of a number of failure conditions to occur in fracture critical members before collapse would require no additional safety factor to be applied to loadings. Structural systems that are considered to be relatively fragile requiring the development of only one failure condition to initiate collapse would be required to have a higher safety factor applied to its loadings and greater levels of quality control during construction.

Snow Load Allowance AS/NZS 1170.3

1. The authors support the recommendations from Section 6 of the NIWA report for further enhancement to the Building Code in relation to snow loading. In particular the snow density parameters and the current reduction factor of 0.7 for snow




loads on roofs with slope less than 10 degrees, as specified in AS/NZS1170.3:2003 should be reviewed as a matter of urgency.

2. It is not known why snow loadings have been allowed a much higher probability of exceedance than earthquake or wind loadings in the current loadings standard AS/NZS 1170.0:2002 (SAA 2002). This should be reviewed with consideration given to making the probabilities of exceedance consistent for all three hazards.

3. The AS/NZS 1170.3:2003 provisions for sub-alpine snow conditions should therefore be thoroughly reviewed in conjunction with an appropriate research programme to better ascertain return periods and snow densities for sub-alpine areas in New Zealand.

Steel Design and Construction Standards NZS 3404

1. A specific reliability calibration of the New Zealand Steel Structures Standard’s for structural performance reviewing what the load combination factors should be used with snow loads in AS/NZS 1170.0:2002.

2. Require inspection during construction of closed section SHS and CHS tubes for compliance with specified wall thickness using ultrasonic testing techniques for fracture critical members in structures of high importance.

3. Specifically require the use of SP quality welds for all welds which constitute part of the main load carrying path, irrespective of stress level, identified for the collapse limit state, whether that is induced by seismic, snow, wind or other superimposed loading conditions. For structures of high importance as set out in C3.2.3 NZS 3404.1:2009 the constructor should have a welding quality management system in place as recommended in C3.2.3. While there are significant limitations to AS/NZS ISO 3834 referenced in C3.2.3, particularly in respect of site welding of structures, it forms a useful starting point for the development of something that could be applied consistently to improve reliability of site welded steel structures.

4. Site fabrication and welding should be supervised and executed by appropriately qualified personnel and companies experienced in steel construction procedures. The steel construction supervisor needs to understand the risks, resources and skills required to achieve the welded construction quality required by the New Zealand Building Code. The draft requirements for Steel Structures Licensed Building Practitioners developed by




industry in conjunction with the Department of Building and Housing should be considered for review and implementation.

5. Where complete penetration welded splices are located in fracture critical members in structures of high importance these should be subject to mandatory 100% UT (ultrasonic or radiographic) examination irrespective of the general provisions of Table 34 NZS 3404.1:2009. The allowance in Note 3 of Table 34 to reduce the amount of UT on the basis of section wall or plate thickness limits in Table 32 should be reviewed. For complete penetration welds in fracture critical members in structures of high importance this reduction of testing should be removed.

6. Review and inspect existing important roof structures with this type of hollow section construction. Particularly where complete penetration site welding has been carried out at collapse critical locations.

Concrete Design and Construction Standards NZS 3101 and NZS 3109

1. A statistical reliability calibration of the New Zealand Concrete Structures Standard review the load combination factors currently used with snow loads in AS/NZS 1170.0:2002 and strength reductions factors for slender column elements.

2. The use of cast-in inserts such as TCMs in fracture critical connections that form part of the collapse limit state structural system (such as were in the connection of Trusses 1 to 9 to the column heads on Grid 5) should be more carefully controlled to ensure reliable performance can be attained in an over-load situation.

New Zealand Building Code and Associated Compliance Documents

No specific changes required, except referencing changes to Standards as outlined above.

Design Review and Construction Monitoring

1. Design reviews to ensure that collapse critical components have been appropriately design and documented, should be mandatory for structures subject to snow loads such as stadiums and other public buildings such as retail stores, libraries and the like.

2. Construction monitoring for steel structures should be set in accordance with the recommendations of NZS 3404.1:2009 which require awareness of the quality control systems of the




steel constructor and the IPENZ: ACENZ guidelines. The purpose being to ensure that the constructor interprets the Design Engineer’s requirements correctly and applies the appropriate level of skill and resources to its execution. This level of monitoring should also ensure that, any irregularities or omissions in the design documentation not previously recognised are able to be identified and rectified. To this end there is a strong preference for construction monitoring to be directed by the Design Engineer.



15 CONCLUSIONS

The collapse of the Stadium Southland after a heavy snowfall on 18 September 2010 has been carefully analysed utilising observations for the Site Examination, the Laboratory Examination and 3D computer analysis.

The likely collapse sequence has been determined and the causes of the collapse identified and reported.

Recommendations for change to the way public buildings subject to snow loadings are designed and constructed in New Zealand have also been made so that another similar collapse may be avoided in future.



16 REFERENCES

Bridge, R. Q. and D. J. Fraser (1987). "Improved G-Factor Method for Evaluating Effective Lengths of Columns." Journal of Structural Engineering 113(6): p.1341-1356.

Dimond. (August 2007). "Dimond Structural Systems Manual." Retrieved 8 October, 2010, from http://www.dimond.co.nz/Architects--Specifiers/Structural-Systems-1152.htm.

Hendrikx, J. (2010). Snow Storm Data Collection: Mobilisation in September 2010. Christchurch, National Institute of Water & Atmospheric Research Ltd.

Hyland, C., W. G. Ferguson, et al. (2007). Assessment of Cyclic Ductile Endurance of Structural Steel Members. Pacific Steel Structures Conference 2007, Wairakei, New Zealand, N.Z Heavy Engineering Research Association (Inc.).

Hyland, C. W. K. (2010). Laboratory Examination and Testing of Components of the Collapsed Stadium Southland Roof DBH101126. Auckland, Hyland Fatigue + Earthquake Engineering.

Leicester, R. H. (1985). Computation of a Safety Index. Civil Engineering Transactions, The Institution of Engineers, Australia: pp.55-61.

Pham, L., R. Q. Bridge, et al. (1986). Calibration of the Proposed Limit States Design Rules for Steel Beams and Columns. Civil Engineering Transactions, The Institution of Engineers, Australia: pp.268-274.

SAA (1995). Structural Steel Welding: Part 1: Welding of Steel Structures Standard AS/NZS 1554.1:1995. Sydney, Standards Australia.

SAA (2000). Mechanical Properties of Fasteners made of Carbon Steel and Alloy Steel Part 1: Bolts, Screws and Studs Standard AS 4291.1:2000 Sydney, Standards Australia.

SAA (2000). Structural Steel Welding: Part 1: Welding of Steel Structures. Sydney, Standards Australia.

SAA (2002). Structural Design Actions : Part 0: General Principles AS/NZS 1170.0:2002, Standards Association of Australia.

SAA (2003). Structural Design Actions : Part 3: Snow and Ice Actions AS/NZS 1170.3:2003, Standards Association of Australia.

SNZ (1992). New Zealand Structural Loadings Standard: Commentary NZS 4203:1992. Wellington, Standards New Zealand.

SNZ (1997). Steel Structures Standard NZS 3404:Part 1:1997. Wellington, Standards New Zealand.


APPENDIX A: STRUCTURAL ANALYSIS AND 3D COMPUTER MODELLING

INTRODUCTION

This Appendix summarises the structural analyses that have been carried out, the assessment of snow loadings and determination of demand to capacity ratios for critical components.

DESIGN SNOW LOADS AND LOAD COMBINATIONS

In this section, as summarised in Table 2 an evaluation and comparison is made of ultimate limit state design snow loads according to the following three documents:

NZS 4203:1992. (SNZ 1992) This was the applicable Standard for design loadings at the time the Stadium was designed in 1999.

AS/NZS 1170.3:2003, (SAA 2003) the Structural Design Actions Standard for snow; and,

AS/NZS 1170.3:2003, the Structural Design Actions Standard, as amended by New Zealand Building Code compliance document B1 with amendment no. 9 effective from December 2008. This sets the current requirements for design loadings for snow in New Zealand.

By way of explanation there are normally two limit states that need to be considered in the design of structures in accordance with the above Loadings Standards (and in accordance with the relevant material design Standards). These are the ‘serviceability limit state’ (SLS) and the ‘ultimate limit state’ (ULS).

The serviceability limit state is concerned with the behaviour of the structure in-service and for which the Standard sets limits on in-service performance parameters such as deflections and vibration.

The ultimate limit state is the condition that ultimately determines whether the stability and strength of the structure is adequate according to the Standard. Quoting clause 2.4.3.1 from NZS 4203:1992, in relation to the ultimate limit which states: “The building as a whole and all its members shall be designed to support the combinations of factored loads and forces in 2.4.3.3 to 2.4.3.6 inclusive.”

In Table 2 the snow on ground loads are shown adjusted by the various factors in the relevant loadings standards to arrive at the roof snow load Su. Su is then factored by the relevant load combination factor, which was 1.2 for NZS 4203:1992. In the current loadings standard AS/NZS 1170.0:2002 (SAA 2002) there is no load combination factor for Su.


APPENDIX A: STRUCTURAL ANALYSIS AND 3D COMPUTER MODELLING continued


Loadings Basis Reference

Snow on

Ground

Sg (kPa)

ULS Design Snow Loads for Roofs

Sr

(kPa)

Ratio Sr/Sr(2003)

NZS 4203:1992

Clause 6.3.2.2(a)

Clause 6.3.2.2(b)

0.30 1.2Su = :

1.2 x 0.264 = 0.32 over entire roof; or

1.2 x 0.33 = 0.40 over Community Courts or Events Courts only (Most relevant for Stadium)

0.8

1.0

AS/NZS 1170.3:2003

0.51 Su = sg x µi

= 0.51 x 0.7 = 0.36

Su (min) = 0.40

1.0

AS/NZS 1170.3:2003

+ B1 amendment 9 from Dec 2008

0.90 Su (min) = sg x µi

= 0.9 x 0.7 = 0.63

1.58

It can be seen from the ratios close to 1.0 in Table 2 that the factored ULS design roof snow load has remained effectively the same from 1992 until December 2008. The ground snow load allowance appears to have increased significantly during that time, however this is because the load factor and the building importance factor were incorporated within the ground snow load directly from 2003. In 2008 the ground snow load was increased by approximately 76% under New Zealand Building Code Compliance Document B1 Amendment 9, after snow induced building collapses in South Canterbury.

Structures designed and built in accordance with the Standards should have a low probability of collapse even when the design capacity ratio reaches the limiting value of 1.0. This is due to the use of combinations of loads and load factors for design, material strength reduction factors and statistical variation of material and member properties. These effects are accounted for in the calibration of load factors and strength reduction factors used in the loadings and materials Standards based on statistical reliability analysis techniques.

Regarding the combination of loads for design, the combination of permanent ‘dead’ load and snow load is required for design according to the 1999 Standard, not the snow load alone. Dead load includes the self-weight of the roof structure and all the other building materials and services that are supported by the roof structure.




The New Zealand Steel Structures Standard NZS 3404 is based for non-seismic design purposes on AS4100. Structures are designed and fabricated using similar materials and loadings standards making the calibration for AS4100 directly relevant to New Zealand steel structures.

The calibration of the Australian Steel Structures Standard AS4100 was performed at the transition from working stress to limit state design in 1986 (Pham, Bridge et al. 1986). This calibration incorporated a reliability study

which recommended the capacity reduction factor for steel members f be set at 0.9, in the Standard.

This was on the basis of a minimum safety index of 2.5 being calculated for failure of a compression strut designed using factored dead load actions alone of 1.25G. This is equivalent to a 1% probability of failure should the factored dead load G be achieved in service (Leicester 1985).

For designs of fully restrained beams based on a factored dead and live load combination of 1.25G + 1.5Q, with the dead load G and live load Q components being equal, as is common for roof structures, the safety index rose to 3.6, which is equivalent to a probability of failure upon attainment of the factored loads of less than 0.1%.

No specific calibration was reported for load combinations including snow load with chords of trusses acting as laterally restrained compression struts. However, as the lower bound safety index for a compression strut of 2.5 was associated with factored dead load it can be conservatively assumed that the safety index would have been greater than 2.5.

The average snow on ground measurement by NIWA the day after the collapse was 0.45 kPa. This was greater than the snow on ground allowance of 0.30 kPa used as a basis to calculate factored design loadings for the roof using NZS 4203:1992 of 0.40 kPa (Table 2). The measured snow on ground load was less than the 1 in 250 year allowance in AS/NZS 1170.3:2003 of 0.51 kPa. In that loadings standard the factored roof snow load was considered to be 70% of the snow on ground load, or 0.36 kPa, but not less than 0.40 kPa for design purposes. If the 70% reduction factor of AS/NZS 1170.3:2003 for roof snow loads was applied to the snow on ground load of 0.45 kPa measured by NIWA then the corresponding roof load would have been 0.32 kPa.

The design roof snow load specified currently in B1 Amendment 9 is 0.63 kPa.

The authors consider that the snow load may have increased from the time of the Stadium collapse to the time of the NIWA Report measurements, as explained elsewhere and as evidenced by the collapse of the nearby Shop Roof the day after the Stadium collapse. The snow load on the Stadium roof at the time of collapse is estimated to have been as low as 0.30 kPa.

The Stadium Southland roof was required however to sustain factored design loading allowing for the combined self-weight of the roof and superimposed snow loadings of 1.2G + 1.2Su equal to 0.80 kPa.




The Failure Analysis found that based on the results of the testing undertaken during the Laboratory Examination, that collapse appeared to have initiated at a combined self-weight and snow loading of G+0.30 kPa or 0.63 kPa.

The roof and ceiling self-weight at collapse was estimated based on the drawings and observations from the Site Examination. An allowance was made for wiring and services. The Trust advised that curtains were hung in the Events Courts area and basketball back boards supported off the Spine Trusses at the time of the collapse. Using the weights supplied this would have added approximately 0.02 kPa onto the roof dead load supported by the Spine trusses, but not sufficient to make their connections critical to initiate the collapse. The curtains and backboards would have had no effect on the dead loads supported by the Community Courts trusses.

A superimposed snow load on the roof of 0.47 kPa should therefore have been able to be sustained with less than a 1% probability of failure initiating.

In conclusion, the load that could have occurred on the roof from the snow, sleet and rain over the whole weekend of 18 September 2010 was high and may have come close to the factored design loadings. Even so it should not have been sufficient to have caused the roof to collapse if the Community Courts trusses and connections of the spine trusses T10 and T11 to the concrete columns had been constructed fully in accordance with the Drawings and relevant Standards.

It would seem appropriate that a statistical reliability calibration study be undertaken to review the appropriateness of current New Zealand snow loadings and members designed using the steel and concrete standards.

STRUCTURAL 3D COMPUTER ANALYSIS

A 3D model of the stadium roof structure was developed in order to assess the various member design actions, so that capacity ratios could be calculated. Reference is made in the following sections of the report to the roof truss, column and grid line numbers from Stadium Structural Engineer’s drawings and as shown on the 3D model in the A3-size Appendix D on page 127.

For this particular roof structure there were three-dimensional effects such as two-way spanning and continuity effects that altered the distribution of load between the various elements of the primary structure. For this reason a three dimensional structural analysis model of the relevant portion of the roof structure was developed to enable the structural actions in each member and connection to be more accurately determined.

A three dimensional linear elastic first order structural analysis was carried out for the portion of the roof structure relevant to the collapse using the structural analysis software program Microstran V8. The following elements were included in the computer model:




Pre-cast concrete columns on Grid 5.

Community Courts roof trusses T1 to T9.

Spine trusses T10 and T11 and supporting concrete columns C13 to C16 on Grid 4 and 4.5.

Steel frame and edge of concrete upper mezzanine floor supporting Community Courts trusses T6 to T9 on Grid 4 between A and E.

Secondary Trusses T16 to T24 above the Events Courts.

Lean-to portal frames above the tiered seating to the Events Courts.

Primary Trusses T12 to T15 above the Events Courts and supporting pre-cast concrete columns C17 to C20 at the northern side of the building beyond Grid 1.

Structural steel and pre-cast concrete wall elements of the control room and upper control room at grid intersection E4. This is to model the effective east –west restraints to bending and buckling of columns C15 and C16. The timber floor and wall framing components of the Control Rooms however were not included in the model.

Pre-cast concrete walls framing in to column C13.

Assumptions and points to note about the structural analysis model are as follows:

The intention was to model the structure as realistically as possible to represent the structure as we understand it was built ie as it stood just prior to the collapse on 18 September 2010.

Care was taken to input the geometry accurately, including all the eccentricities between the various roof trusses and their connections to supporting elements, for example the eccentricities in two directions of the connections of Trusses T11 and T15 to pre-cast column C15. Eccentricities such as these impart significant secondary moments into the structure and these secondary moments were able to be output directly from the computer analysis.

Two load cases were analysed. The first was the ultimate limit state design load combination 1.2G+1.2Su from NZS 4203:1992, with Su=0.33 kPa ie the critical design load combination that applied in 1999 when the structure was designed and constructed. The second load case was the estimated load combination at initiation of collapse ie G+0.30 kPa.


APPENDIX B: EVALUATION OF CRITICAL STRUCTURAL COMPONENTS

METHODOLOGY

Critical structural elements and details of the building have been evaluated in terms of design basis and failure basis.

The design basis evaluation used loadings, component capacities and safety factors derived from the Standards and manufacturer product information.

The failure basis evaluation involved a detailed review of site observations and photos in conjunction with calculations to assess collapse process, actions on, and capacities for each component. The calculations for the failure analysis use the actual assessed loadings, without load factors, and realistic values for material properties leading to failure capacities.

ROOF PURLINS

Design Basis Evaluation of the Purlins

The purlins over the Community Courts between Grids 4 and 5 were spaced at 1.74 m centres and were simply supported DHS 250/18 with 2 braces in the longer span end bays and DHS 250/15 with two braces in the interior bays.

There was conflicting information shown on the Drawings regarding purlin thicknesses over the Community Courts. The purlins were noted as DHS250/15 on drawing S72 and DHS250/13 on drawing S16A. However, the 1999 Review Report, together with advice from the Insurer’s Engineer who was also investigating the collapse, confirmed the purlins over the end bays of the Community Courts as 250/18 and other purlins over the Community Courts as 250/15. The Insurer’s Engineer reported that micrometer measurements of purlin material extracted from around bolts at cleats on various trusses confirmed the end bay purlins to be 250/18.

According to the Drawings all of the relevant roof purlins had two proprietary bracing channels per bay. Taking into account the as-built purlin thicknesses outlined above we have reviewed the design of purlins in accordance with NZS 4203:1992 and the manufacturer’s Design Manual as follows:

The weight or dead load of the roof and ceiling supported by the purlins including roofing, bracing channels, insulation, wire mesh, building paper and suspended ceiling was calculated to be 0.22 kPa.

The uniform load on purlins due to self-weight and roof dead load was calculated to be 0.26 kPa or 0.46 kN/m.


APPENDIX B: EVALUATION OF CRITICAL STRUCTURAL COMPONENTS continued


Purlin Location

DHS Purlin Spec

Purlin Span (m)

Purlin Spacing (m)

ULS Load 1.2G+ 1.2Su Per NZS 4203:1992 (kN/m)

Bending Capacity from DHS Purlin Design Chart (kN/m)

Ratio of ULS Load / Bending Capacity

Community Courts End Bays

250/18 10.6 1.74 1.22 1.47 0.83

Community Courts Internal Bays

250/15 9.2 1.74 1.22 1.63 0.75

Events Court

250/15 7.2 1.54 1.08 2.53 0.43

Table 3 that the purlins above the Community Courts and above the Events Courts did comply with the strength requirements of the relevant 1999 design Standards. The highest ratio of ULS load to bending strength capacity is 0.83 for the Community Courts end bay purlins.

The 1999 Review Report stated “It must be noted that as the building is now almost complete some elements which do not meet serviceability code recommendations may have to be accepted and the risks managed in some other way. The ultimate strength of these particular elements have been confirmed as adequate and although not really satisfactory are not considered unsafe.” The items identified in the 1999 Review Report as not meeting code recommendations included “purlin serviceability wind deflections and purlin connections in end bays over the Community Courts.”

According to the Commentary to NZS 4203, which was an informative rather than normative part of the standard, the deflection of purlins should be considered so as to avoid the possibility of excessive ponding effects under the dead load and serviceability snow load combination.

Quoting from Note 8 to Table C2.4.1 in the commentary to NZS 4203: “Flat roofs are susceptible to ponding when drainage is precluded. Where parapets are used, or with wide span flexible roofs, the weight of water and snow may induce instability as a result of ponding deflections.”

The Community Courts roof could be interpreted to not have been a “flat roof” as defined by NZS 4203:1992, however we have checked the serviceability deflections of the purlins in any case.




Under NZS 4203:1992 the appropriate serviceability snow load combination to check deflections is:

G + 0.5Su = 0.26 + 0.5 x 0.33 = 0.43 kPa

Deflections of the purlins under this load combination are calculated to be as shown in Table 4.




Purlin Location

DHS Purlin Spec

Purlin Span (m)

Purlin Spacing (m)

SLS Load G & 0.5Su Per NZS 4203:1992 (kN/m)

Purlin Deflection (mm)

Purlin Deflection (in relation to span)

Community Courts End Bays

250/18 10.6 1.74 0.74 75 Span / 141

Community Courts Internal Bays

250/15 9.2 1.74 0.74 51 Span / 180

Events Court 250/15 7.2 1.54 0.65 17 Span / 423

From Table 4 the maximum purlin deflection of 75mm in the end bays above the Community Courts area was significant, however it has been calculated that this deflection would not have caused ponding. Ponding is prevented by the 5 degree cross-fall of the Community Courts roof in the orthogonal direction.

Failure Basis Evaluation of the Purlins

The damaged truss T1 end bay purlins had been removed by the time of the authors Site Examination and had been disposed of according to the Insurer’s Engineer.




There was no evidence of purlins failing at their mid-span peak bending locations which is evidence that damage observed to the purlins in the end bay had occurred as a consequence of collapse. In general the purlins and roofing lay relatively flat and undeformed on the floor after the collapse.

At the western side of truss T8 two bays of purlins and roof, together with truss T9, had fallen as a unit to the floor swinging in eastwards approximately 9m as it fell Figure 15). The eastern movement of truss T9 is believed to have been caused by the temporary restraint of the purlins connecting to truss T8 before those purlins became detached from truss T8.

At the eastern end bay the purlins had bent at around 1.5 m from truss T1 at the southern end of the truss. The purlin ends had pulled away from the truss cleats in this area and were pointing skywards (Figure 14). This may have been due to the way the truss and roof fell, constrained by the eastern concrete wall.

The DHS 250/18 end bay purlins spanned 10.58m simply supported and had a nominal uniform load capacity Wbx = 1.47 kN/m based upon the manufacturer’s design manual Table 2.3.7 May 2004 (Dimond).

To exceed the nominal capacity of the end bay purlins a roof snow load of 1.01 kN/m or 0.56 kPa would have been required. This is greater than the 0.30 kPa snow load estimated to have caused the collapse of the roof, so




failure of the purlins is not considered to have occurred prior to the roof collapse occurring.

The DHS 250/15 interior purlins spanned 9.18m simply supported and had a nominal uniform load capacity Wbx = 1.63 kN/m based upon the manufacturer’s design capacity tables.

To exceed the nominal capacity of the interior purlins a snow load of 1.17 kN/m or 0.65 kPa would have been required.

The NIWA Report stated the snow on the ground adjacent to the stadium the day after the roof collapse was sg = 0.45kPa.

Therefore it seems clear that the purlin nominal load capacities were not exceeded.

Discussion of Purlin Performance

The analyses and the condition of the purlins after collapse show that the purlins over the Community Courts were not loaded beyond their nominal bending capacity prior to the collapse developing.

The photos of purlins between truss T1 and the eastern Grid G wall however showed a different mode of failure and raised questions as to whether the end bay purlins in this area had been upgraded to the stronger DHS 250/18 purlins.

Measurements by the Insurer’s Engineer of purlin remnants are evidence that DHS 250/18 purlins had been used. Those purlins may have broken their backs at the ceiling up stand to the skylight as the roof and truss T1 hit the floor.

The collapse of the light framed walls on top of the pre-cast concrete wall panels at the western and eastern ends of the community court area also occurred with the collapse of the roof.

Purlins had in a number of cases pulled away from rafter trusses by tearing around bolt holes, which is thought to have been caused as the purlins tried to pull the roof westwards towards trusses T6 to T9 as the collapse of the Community Courts trusses T1 to T5 developed.




SOUTH WALL COLUMN HEADS CONNECTIONS

Design Basis Evaluation of South Wall Column Heads Connections

The Community Courts trusses had originally been detailed for moment-fixity to the south wall column heads as seen in Figure 20. However it was noted in the 1999 Review Report that some cracking of the column heads had occurred during construction.

Remedial work specified in the Drawings following the 1999 Review Report was to first remove all the bolts connecting the trusses to the column heads, to make good fixings to the top of the column only and to repair the cracks to the concrete columns using epoxy grout injection. The intention was then to rely on the columns to support axial loads only.

The design capacity of the 6 TCM inserts in the arrangement shown on the

Drawings has been calculated to befVb= 170 kN using NZS 3101:2006.

The resultant bending capacity provided by the connection to reduce mid

span bending would have been as little as fMb=183 kNm as the trusses rotated and engaged with the north faces of the south wall Grid 5 columns

This is small in relation to the mid-span bending capacity of the trusses. As the bending capacity of the column head connection was limited by a brittle mode of failure it may have been exceeded earlier in the life of the Stadium.

This meant that the south wall column heads therefore had little ability to maintain connection between the column and the trusses as the collapse of the Community Courts trusses progressed.

Failure Basis Evaluation of South Wall Column Heads Connections

The heads of columns C3, C7, C8, C9, C10, C11 that supported trusses T1 to T6 and T9 sheared off in a similar manner as described in the discussion of truss T1 (Figure 21). Similar cracking was observed in the top of column C4 supporting truss T8 which did not collapse.

The 6 cast-in TCM 24 anchors pulled a wedge of concrete off the top of the columns. The fracture surface was found to run below the base of the anchors. There was no transverse confining reinforcement through the failure surfaces of the connectors. The failure surface was consistent with shearing of un-reinforced concrete as the top plate of the trusses pulled inwards.

A number of the connectors on the top of the column were missing bolts which appeared not to have been installed.

The top chords of the Community Courts trusses adjacent to the column heads showed no signs of yielding at the column. This showed that the bending resistance developed at the connection of the truss to the concrete column was below the flexural capacity of the truss. The mode of failure of the attachment of the columns to the trusses would have been brittle and




sudden. However the trusses would have remained in place due to friction until the westward pull on the roof due to the increasing vertical displacements near mid-span, overcame the remaining resistance in the connections at the column heads.

Discussion of South Wall Column Heads Connections Performance

During the Site Examination cracking could be seen in the head of column C4 supporting truss T8 which did not collapse. This was consistent with the calculated moment capacity of the connections for the trusses being limited by shearing failure of the columns heads at loads well below what would be required to fail the trusses themselves.

The non-ductile nature of these connections and the use of TCM inserts in them is considered to be inappropriate by the authors for a location such as this.

The inability of the column head connections to prevent the trusses being pulled from their supports contributed to the progressive nature of the collapse.

COMMUNITY COURTS ROOF TRUSSES

Design Basis Evaluation of Community Courts Trusses T1 to T9

The original design for the Community Courts roof trusses, as shown on the drawings submitted to Council for the Building Consent had 125x9 SHS chords throughout (Figure 5). The remedial works drawing 97139 indicated that some trusses had been fabricated and installed at that point using 125x6 SHS chords.

The end fixity to the Grid 5 columns was found not to be adequate during construction. Consequently, following the 1999 Review Report, the connections to the tops of the Grid 5 columns were to be released from providing moment fixity, and the roof trusses strengthened to span simply supported instead.

From the Site Examination of truss T9 and advice from the Insurer’s Engineer during the deconstruction of trusses T6 to T8 after the roof collapse, trusses T6 to T9 had all been fabricated with 125x9mm SHS chords.

The Laboratory Examination found 5 mm wall thickness SHS chord in the northern half of the bottom chord of truss T4, rather than 6 mm wall thickness found elsewhere in trusses T1 to T4.

Wall thicknesses of 3 mm rather than the specified 4 mm thicknesses in the RHS diagonal ties were found in parts of trusses T1 and T4.

Trusses T1 and T9 carried more load than the other Community Courts roof trusses. This was because the end bay roof purlins had a longer span than those in the internal bays.




A summary of design actions from the 3D structural model for Community Courts roof trusses T1, T4 and T9 chords, is shown in Table 5. These actions take into account the as-built support conditions. The relevant chord sections with and without side plating are listed. The axial section design capacities in the Table were calculated based on the nominal properties for Grade C350 SHS sections and Grade 300 strengthening plates.

Trusses T1 to T3 were fabricated with 125 x 6 mm SHS chords and with an overall depth of 1200mm. The design mid-span moment capacity of the trusses assuming Grade C350 SHS chords and without the 100 x 10 FL

side plates was fM= 924 kNm. With correctly installed Grade G300 side

plates this would have increased tofM= 1505 kNm.

Truss No

Chord Section

Chord Design

Axial Action at Midspan

NZS 4203:1992

1.2G+1.2Su (Su =

0.33kPa)

N* (kN)

Chord Design

Compression Capacity

fNc (kN)

Design Capacity

Ratio

N*/fNc

Chord Failure Axial

Action at

Midspan

G + 0.30 kPa

Nf* (kN)

Chord Actual

Capacity from Tests

Nf (kN)

Failure Capacity

Ratio

Nf*/Nf

T1 125x6 SHS

1186 861 1.38

T1 125x6 SHS +

2 x100x10 FLATS

1186 1401 0.85 941 941 1.0

T4 125x5 SHS

1170 729 1.60

T4 125x5 SHS +

2x100x10 FLATS

1170 1269 0.92

T9 125x9 SHS

1293 1230 1.05

T9 125x9 SHS +

2x 100x10 FLATS

1293 1770 0.73

Tensile testing of the truss chords and the strengthening plates showed them to have tensile yield stress properties significantly greater than the nominal design properties meaning they could have sustained bending




demands greater than these design capacities if they were fabricated in accordance with the Drawings and Standards.

In Table 5 it can be seen that, without side plates, the SHS axial load ratios varied from 1.05 to 1.60 ie greater than 1.0 and would be unacceptable. This shows why the remedial works to the Community Courts roof trusses were necessary.

With the specified side plates added the design axial capacity ratios varied from 0.73 to 0.92 ie less than 1.0 which would have been acceptable according to the Standards applicable in 1999 at the time of construction.

The trusses developed failure hinges at mid-span due to top chord compression tearing failure initiating from 3-sided end bearing at the splice plates. The bottom chord SHS weld failed also with some localised yielding of the strengthening plates due to combined axial tension and local bending as the chord rotated after top chord failure progressed.

The reduction in truss bending capacity due to these localised failure conditions occurring was significant. Truss chord assemblies were tested in compression during the Laboratory Examination (Hyland 2010). The results are summarised on page 125 of this report.

The fully restrained axial capacity of the top chord without the side plates, with 4-sided end bearing assuming steel with yield stress of 350 MPa (consistent with lower bound grade C350 AS/NZS 1163 SHS sections) was Ns = 861 kN.

Whereas for 3-sided bearing this reduced to Ns =693 kN. The theoretical bending capacity of the section reduced from Ms= 42 kNm for 4-sided to Ms= 24.9 kNm for 3-sided bearing.

Therefore the 3-sided bearing effect alone was estimated by calculation to reduce the nominal section axial capacity of the chord by around 30%.

Tensile testing of the 125x6 SHS chord steel found the yield stress measured as the 0.2% proof stress to be on average 436 MPa which is higher than the 350 MPa nominal grade stress used for design.

Compression testing of salvaged chord elements, found the axial capacity to be as low as 941 kN when a preset 8 mm offset of the bearing plate from one surface was used. Side plates were kept discontinuous over the splice plate. Failure initiated with compression induced tearing at the splice plate and localised buckling at the strut ends (Figure 49 in Appendix C).

A superimposed snow load of 0.30 kPa was determined to have been sufficient to cause axial compression failure of the top chord of truss T1 and truss collapse. This was based on actions derived from the 3D computer analysis and the chord strut compression tested capacity of 941 kN (Table 5).

The top chords punched through their abutting chord sections near mid-span at the location of the chord splice packing plates. These packing




plates had been inserted as part of the strengthening works. This tore the top of the chord along the line of the strengthening side plate. In the process the strengthening side plate on the eastern side of the chord had been peeled off by around 250 mm (Figure 16).

The top chord packer plates on the south side of the joint were installed with only 3-sided end bearing to the chords and did not have full penetration welds connecting them to the chord as specified in note 5 of the modifications drawing (Figure 52 in Appendix D). In fact the top weld had been omitted completely.

Failure Basis Evaluation of Community Courts Trusses

Common Features of Trusses T1 to T3 Failures

The as-built chord capacity weakened by the splice configurations was found from the Laboratory Examination to be 67% of the calculated design capacity. It was 21% less than the required NZS 4203:1992 design actions (Table 5).

Trusses T1, T2 and T3 collapsed in a similar way, resting in a similar manner on the ground. All these trusses fell directly to the floor of the Community Courts after pulling in and away from the Grid 5 south wall column heads and breaking free of spine truss T10 on Grid 4.

These trusses settled with a 4.5 degree angle from the horizontal, back to Grid 5. The total permanent vertical displacement at mid span relative to a line drawn from the truss ends was approximately 1.5 m scaled from photos. This meant a permanent northwards pull-in of approximately 0.12 m occurred at the supports. This would not have been enough pull-in by itself to have caused the trusses to come off the south wall concrete column heads at Grid 5, as the seating plate on the trusses was 750 mm long. Therefore trusses T1 to T3 must have come off the Grid 5 columns as spine truss T10 on Grid 4 fell to the ground.

The pull-in action at the supports from the sagging trusses was calculated to be in the order of 50 to 100 kN. This was limited by the shearing reaction provided by friction of the trusses on the concrete columns at Grid 5.

The bottom chord cleat that connected truss T1 to T10 fractured in tension. This cleat had been site welded to truss T1 and to truss T10 thereby becoming effectively a fully welded (and not bolted) connection.

The M20 threaded rods used to connect all the other chords of trusses T1 to T3 to cleats on truss T10 had sheared off.

All the trusses T1 through to T9 were modified following the 1999 Review Report (Figure 52 in Appendix D). The strengthening work apparently occurred on site after the roofing had been installed which would have meant the skylights should have been removed to effect the remedial works (Figure 5).




The top chords of the Community Courts trusses were restrained laterally by the eccentric restraint of the purlins attached to the vertical struts of the trusses and the tension in the bottom chord (Figure 5).

While a relatively complex arrangement calculations show it to be adequate and there is no evidence that this arrangement of lateral restraint led to the collapse condition that occurred.




Observations of Truss T1 Failure

The bottom chord of truss T1 fractured in the mid-span splice weld. This weld was likely to have been made on site during construction, to connect shop fabricated portions of the trusses together prior to erection. There was some 45o shearing of the weld consistent with tensile failure that then ran perpendicular to the surface. The internal plug plate may have been over-sized. There appeared to be a lack of fusion of the weld to the internal plug plate. The weld had a very rough surface profile (Figure 17).

Confirmation of the extent of welding was made during the Laboratory Examination phase and found to be an average thickness of 2.6 mm Design Throat Thickness (DTT) or 43% of that required to achieve a full penetration weld through the full wall thickness of 6 mm.

There was very little yielding evident adjacent to the fractured weld in the bottom chord due to the white paint still being intact either side. It was also confirmed by tensile tests of the chord at various locations in the trusses that no obvious yielding occurred in the truss chords away from the welds that fractured. The strengthening plates either side of the fractured chord had elongated significantly. The stitch welds between the SHS and the strengthening plates had sheared off around 375 mm either side of the chord fracture.

The stitch welds between the strengthening plate and the chord of the truss were small on all the trusses observed, with high levels of porosity as can be seen on the stitch weld photo from truss T4 (Figure 18). They appear to have been welded on site through the paint on the truss chord, without this being removed as it should have been, prior to welding.

The strengthening plates had not been end to end butt welded as specified in note 6 of the structural steelwork modifications drawing (Figure 52 in Appendix D). No specification for stitch welding of the plates was shown on that drawing.




One of the diagonal roof brace rod cleats had pulled out of the truss vertical strut adjacent to the failure location. There was some bending deformation in the strut at the rod connection location. The tie rod cleats did not pass through the struts but were face welded to the struts on the cold formed corners. These had pulled away as a unit in a number of locations on the roof trusses (Figure 19).

The south side of the top chord bent downwards as it punched through the abutting chord (Figure 19).

The paint scrape marks on column C11 from truss T1 showed that the south end of the truss pulled off northwards in line with its axis and then dropped directly to the floor below (Figure 20). The truss top plate had sheared the top off the column with minimal damage to the TCM inserts. An angled slope to the top of the column either side of the centre formed. There was very little effective confinement to the concrete around the TCM anchors to prevent the brittle failure and sudden loss of shearing capacity (Figure 20).

The north end of truss T1 had been connected to the spine truss T10. The top hole in the truss T1 chord had a 20 mm diameter mild steel threaded rod passing through it connecting either side to 2 cleat plates. The hole in the truss had ovalled horizontally and the rod had sheared off.

The bottom chord support of truss T1 onto truss T10 had a threaded rod bolted through the slotted cleat plates. However the cleat plates had been welded both to truss T1 and to truss T10 preventing movement. The cleats fractured in combined tension and shear. The cleats had been made up of a series of packer plates that had rough gas cutting marks obvious on the surfaces. These had been welded together without any backing strips and without preparation of the joining plates. The welds had little penetration below the outside surface of the cleat. Shiny cleavage or inter-granular faceted fracture surfaces were apparent on the weld showing that the weld fracture occurred suddenly (Figure 17).

All the trusses T1 to T5 broke away from spine truss T10. The vertical struts on truss T10 were seen on site to have bent outwards with maximum permanent deformation at the location of the bottom chord of the Community Courts trusses (Figure 23). The slotted bottom cleats of




trusses T2 to T5 remained on the supporting struts of truss T10 but the bottom bolts had sheared off.

Discussion of Truss T1 Failure

Both the top chord and bottom chord failures originated at locations where the as-built steelwork did not conform to the requirements of the Drawings and Standards.

Based on the results of the compression strut testing on page 125, the top chord failed prior to it reaching its design capacity.

Laboratory tensile testing of samples from the top chord indicated that it did not develop compressive stresses at or above the yield stress of the steel except locally at the butt splice failure.

Steel strains of around 5% were found during tensile testing to initiate cracking and flaking of this paint on the steel surface. Zones on the trusses that had suffered 5% strain or more can therefore be identified by the grey painted surfaces that have been exposed by cracking of the white top coat paint.

The bottom chord SHS tube fractured after the top chord failed and the truss rotated significantly at that location. Due to the connection of the strengthening plates to the SHS chord there must have been compatibility of strain in the chord and strengthening plates up until the chord weld fractured. A combination of axial tension and bending is considered to have led to the fracture of the bottom chord weld once the top chord failed in compression.

This is confirmed by calculation of the tension only capacity of the splice at the weld. This used the measured average weld thickness DTT of 2.6 mm and the ultimate tensile strength of the weld metal equal to that of the chord at 496 MPa.

The side strengthening plate yield stress was 340 MPa as measured during testing.

The fracture capacity was calculated to be 1289 kN which would require a superimposed load of 0.54 kPa to develop.

Collapse of the truss by top chord compression failure at 0.30 kPa was therefore the governing limit state for truss T1. The strengthening plates on the bottom chord in this case would then have held the bottom chord together until the truss fell to the floor.

The tensile failure at the weld in the bottom chord is indicative of incomplete weld penetration and fusion. The thickness of the plug plate used as a backing plate to the complete penetration weld would have affected the cooling rate of the weld and may have led to poor fusion. Where the weld was undertaken in the over head and vertical positions the welding requires a far greater level of skill than when laid in a down hand position.




The weld would not have passed a visual examination for SP quality welds in accordance with NZS 3404:1997 (SNZ 1997) and its referenced welding standard AS/NZS 1554.1:1995(SAA 1995; SAA 2000).

The small size and high levels of porosity in the stitch welds connecting the side strengthening plates to the top chords and lack of end to end butt welding of the plates meant that there was little development of the strengthening plates into the chord. They would easily have been peeled off by the applied actions. The strength of the chords therefore would have degraded suddenly as the strengthening plates disconnected from the top chord.

The horizontal distortion of the top hole in the truss T1 connection to truss T10 indicates that the top bolt was pulling away from the Truss 10 connection prior to the bolt shearing (Figure 23). This would be due to a combination of flexural hogging and tensile pull-in as truss T1 hinged at mid-span and started to displace downwards severely.

The welding of the bottom chord cleat appears to have occurred during the structural modifications (Figure 22). The tensile and/or shear failure of the bottom chord cleats would likely have been initiated as the truss displaced downwards and inwards as failure hinging developed at mid span of the truss. Rotation of spine truss T10 before or during its collapse would have also led to the development of second order bending effects that may have reversed initial hogging actions at the support. Once the bottom cleat fractured the top chord bolt would have sheared off.





Truss T2 showed a similar pattern of failure as Truss 3 with top chord peeling and crushing at the mid-span packer splice (Figure 24). The top weld was omitted and fit up was such that only 3-sided end bearing occurred at the packer plate, reducing the end bearing capacity of the top chord.

The packer plate had only minimal amounts of weld connecting it to the abutting chords.

No complete penetration weld had been made as specified.

No backing plates were found as required to pre-qualify such a weld in accordance with AS/NZS 1554.1:1995 (SAA 1995) had been installed.

The bottom chord side plates had yielded locally adjacent to the fractured SHS welded splice location.

The white truss paint was still intact except locally at the mid-span failure locations.


The authors conclude that localised top chord peeling and crushing due to 3-sided bearing at the mid span splice initiated mid span failure hinge development in the truss.

The stitch welds connecting the side plates to the top chord side plates fractured as the chord bulged due to the compressive distortion in the chord. As the top chord deformed in compression it was forced downwards by its eccentricity, resulting in deformation of the diagonal tie in that bay (Figure 24).

The truss did not develop stresses at or beyond yield stress outside the welds as there was no obvious distress to the white paint and as found from tensile testing of the chords at various locations along the trusses T1 and T4 undertaken during the Laboratory Examination.

Failure was therefore due to non-conformance of the splices and the strengthening plates with the Drawings and Standards.





The top chord of truss T3 had pushed down and below the abutting chord at the mid-span splice location. The splice plate had very little weld to the abutting chord and disconnected. The diagonal tie RHS in the same bay bent under combined axial and bending actions (Figure 25).

The side plates to the top chord were not continuous over the splice joint as specified in the Drawings. There was some localised yielding of the side plates at the bottom chord node point where the SHS welded splice had fractured.


The stitch welds connecting the side plates to the top chord side plates fractured as the chord bulged due to the compressive distortion in the chord. As the top chord deformed in compression it was forced downwards by its eccentricity also resulting in deformation of the diagonal tie in that bay (Figure 25).

The localised yielding of the bottom chord side plates was consistent with fracture of the bottom chord weld having occurred first followed by localised deformation as the truss displaced downwards.

Failure at mid-span was therefore considered to have initiated by compression failure at the top chord mid span splice. It appears that the packing plates at this location only had bearing on the two vertical side and bottom edge of the chord, leading to localised crushing below its design axial section capacity.





Trusses T4 and T5 fell differently to trusses T1 to T3 as can be seen in the aerial photo taken the following day (Figure 2). Whereas trusses T1 to T3 fell as a unit straight down, trusses T4 and T5 pulled westward at their southern ends by around 1.7 m and remained attached by purlins to truss T6 (Figure 26 and Figure 29).

The truss section north of mid-span was relatively straight and in-line with its supports. From south of the mid-span Truss 4 kinked sharply towards truss T5.

There was localised buckling in the bottom chord node adjacent to that The purlins and roof had fully detached from truss T4 and displaced 1.7 m towards truss T5 up to the quarter point approximately 8 m from the south wall at Grid 5, where the truss was also displaced with the roof and purlins (Figure 28).

The top chord had fractured at the southern quarter point where it was found during the Laboratory Examination that no splice plate had been installed during the remedial works (Figure 27).

location. There was no end bearing packer plate or backing strip to the weld in this location, and its failure was examined in more detail during the Laboratory Examination.

The diagonal roof plane bracing had pulled out of the vertical strut on the eastern side.

The SHS bottom chord of this truss from mid-span to Grid 5 were subsequently found during the Laboratory Examination to have a wall thickness of only 4.88 mm (nominally 5mm) compared to the expected 6 mm, as found in T1 to T3. The diagonal tie members were also found to be nominally 3 mm wall thickness instead of the 4 mm specified on Eng Drawing S86.




Fracture had occurred in the mid span weld to the bottom chord and the side plates had yielded as evidenced by the exposed grey primer. The side plates had not fully detached. The development of failure at mid-span was similar to what occurred in trusses T1 to T3 but less advanced.

There was what was found to be a typical lack of fit-up of the end bearing packer inserted in the top chord at mid-span during the modifications. There was bulging to the underside of the top chord indicating compressive yielding had occurred. Tearing of the top chord was starting to initiate as the packer plate starts to cut into the corner of the abutting chord (Figure 30).

The stitch welds were very small, had very high levels of porosity and are non-compliant with the welding standard AS/NZS 1554.1. They appeared to have been laid through the paint on the truss chord.




On the east side the side plate extended past the SHS welded splice. However on the west side the side plate stopped either side of the SHS welded splice.

The truss had broken away from its fixings to spine truss T10. The connecting threaded rods bolts had sheared off.


An explanation for truss T4 falling in a different way to trusses T1 to T3 is the effect of trusses T6 to T8 which remained in place, pulling the roof and trusses westwards towards themselves as they fell, due to the roof purlins swinging down in an arc about truss T6. A similar thing happened to truss T9 which swung eastwards as it fell. At T4 the purlins pulled away from the trusses in places.

The welding of the roof plane bracing cleats to the rolled corners of RHS members is not generally recommended as these are highly cold worked zones that are susceptible to cracking when welded. There were a number of instances where the cleats broke out the corners of these RHS members.

The tie back effect from the purlins connected into trusses T6 to T8 may be the reason for the reduced development of compression and tension chord failure conditions in trusses T4 and T5 relative to trusses T1 to T3.

The substitution of lower wall thickness products in the Community Courts trusses during construction is a cause for concern. The reduced section sizes would have increased deflections of the truss and reduced their as-drawn design capacity.





Truss T5 had three broken top chord splice locations (Figure 31). The break at the southern quarter point was also at the point where the truss had kicked west by around 1.8 m. The rest of the truss appeared to lie in alignment with the supports, however the truss was tilted on its side with the top chord laying eastwards.

The quarter point failure appeared similar to that in truss T4. It was not clear whether there was also bottom chord failure at that location as in Truss 4.


The eastward tilt of truss T5 is consistent with it having rotated during its fall and swung in an arc about its purlin attachment to truss T6. Otherwise T5 had similar issues as T4.

Observations of Trusses T6 to T8 Performance

Trusses T6 to T8 did not collapse. However the column heads at Columns C4, C5 and C6 were severely cracked indicating loss of shear strength restraining the trusses in line with the top chords.

During deconstruction the Insurer’s Engineer reported that these trusses had 9 mm wall thickness chords rather than 6 mm. Under controlled demolition arranged by the Insurer’s Engineer the trusses T6 to T8 were found to fail suddenly at the mid-span bottom chord welded splice in a manner similar to truss T9.




Discussion of Trusses T6 to T8 Performance

Trusses T6 to T8 probably did not collapse as the SHS chords in those trusses had a 9 mm wall thickness. The imposed loads on them were also less than those on truss T9 due to their reduced tributary area compared to the end bay truss T9.


Truss T9 fell with the south end landing on the floor displaced eastwards towards truss T8 by around 9m (Figure 33). Scrape marks on the south wall show the arc the end of the truss moved in as it fell. The north end of the truss remained held up to some extent on the northern wall at Grid 4. The bottom chord sections displaced from each other fracturing the stitch welds in the process and the side plates detached from the northern side of the bottom chord. The side plates had yielded over a length of approximately 2050 mm evidenced by the loss of white paint in that length and tensile testing performed during the Laboratory Examination (Figure 36).

The surface profile of the splice and stitch welds made on site were almost without exception small and full of porosity. The welds did not comply with the visual examination requirements for Category SP (Structural Purpose) quality welds in accordance with the welding standard AS/NZS 1554.1, as would be expected for welding of these trusses.

A heavy internal plug plate was evident on the northern side of the joint. Examination of this in the Laboratory Examination found that the inside faces of the SHS chord had been bevelled, perhaps to enable easier fit up on site during initial erection splicing.

This bevelling compromised the effect of the plug plate as a backing plate and also reduced the maximum weld thickness achievable to be less than the wall thickness of the section (Figure 32 and Figure 43).

The welds to the top chord packer plates were minimal and would have been difficult to lay correctly on site. In this case the top chord packer plate was welded on all four sides on the north face of the splice. But not on the south face. 4-sided end bearing has occurred at the top chord and no bulging or peeling of the top chord side plates had occurred as happened to the trusses T1 to T4 with 6mm wall thickness top chords.

The top weld was not completed at the quarter point of the truss at the south end. No packer plate or backing strips were observed in the splice, making the weld non-compliant in terms of AS/NZS 1554.1 pre-qualified complete penetration welding requirements.

Measurements during the Laboratory Examination of the mid-span bottom chord splices of truss T9 found the butt weld thickness (DTT) of truss T9 to be 3.1 mm rather than the 9.4 mm required for full section thickness (Figure 43).




Fracture mechanics based calculations were made to assess the tensile fracture capacity using actual measured thickness (DTT) of the chord weld and the measured tensile properties of the side plates and chord found in

the laboratory testing. An assumed fracture toughness of 94

based on measurements of G300+ steel (Hyland, Ferguson et al. 2007) was used in the absence of test data. The calculations show that the weld would develop an unstable running fracture at an axial tensile stress of 206 MPa, equivalent to an axial tension action of 1262 kN or superimposed load of 0.46 kPa. This is higher than that calculated from compression failure of truss T1.

This indicates that the collapse of truss T9 may have occurred after initial fracture of the top chord mid-span splice during uplift from sudden build up of pressure to the underside of the roof as trusses T1 to T5 collapsed. This same pressure was witnessed to have blown out the doors and walls at the east end of the Community Courts.

The mid-span top chord was found on site to have broken apart at the packer plates (Figure 34). No weld preparation had been made at the chord splice with the packer plates, or between the packer plates as




necessary for a pre-qualified weld complying with the provisions of the Structural Steel welding standard AS/NZS 1554.1.

There was less than 1mm weld thickness connecting the packer plate and the end of the top chord (Figure 35).

The vertical strut to which the diagonal roof plane bracing was attached at mid span of the truss had broken away from the top and bottom chords. The top weld was not completed at the quarter point of the truss at the south end. No end bearing packer plate or backing strips were observed in that splice (Figure 36).


Rotation of the post support at Grid 4 would have caused a second order effect to develop a positive bending moment at the bottom chord connection due to the eccentricity of the vertical support reaction from the end of the truss. This would have combined with the pull in effect as the truss displaced at mid span after the chord weld fractured but while still held by the side plates as they elongated. The fracture of the bottom chord attachment to the post at Grid 4 may have led to the final debonding of the side plates and collapse of truss T9.

Bevelling of the internal faces of the chord walls at the splice left a significant gap between the chord wall and the backing plate that would have prevented a reliable full penetration weld being made. The throat thickness DTT of the weld that resulted was much less than the wall thickness and so had significantly less strength than the chord it was joining. The fundamental principle of complete penetration welding of structural steel welding is that the weld must not be weaker than the section it is joining.

It seems likely that this weld was made on site prior to erection to join the two ends of the truss together. As with all the welds that have failed they would not have complied with the visual scanning criteria of AS/NZS 1554.1.

The use of thick packer plates would have required the effective use of pre-heat. Backing strips would have been required to comply with normal pre-




qualified welding procedures used in structures in accordance with AS/NZS 1554.1:1995 (SAA 1995).




CONNECTIONS OF COMMUNITY COURT TRUSSES T1 TO T9 AT GRID 4

Design Basis Evaluation of T1 to T9 Connections to Grid 4

A summary of the output from the structural analysis computer model for the Community Courts roof truss connections to Grid 4 supports is shown in Table 6. The structural model took into account the as-built support conditions observed on site during the site visit.

In Table 6 the shear action on the threaded rod connections has been calculated as the average shear in the top and bottom rods for each truss. The 3D linear elastic analysis gave different shear values for top and bottom rods, however it is accepted in structural engineering practice that these can be considered to redistribute to an average value for each truss chord, which is what is shown.

Where the bottom cleat was welded to the truss chord of T1 and T9 the shear may reasonably be considered for design purposes to have been totally resisted by the bottom cleat. In this case the top chord bolt could have been considered to have no shear demand placed upon it. However for completeness a range of shear actions have been shown for shear actions V for T1 and T9.

The higher axial actions on the truss T1 connections are due to tensile actions developed in the top chord of the truss from the bottom chord being fixed. As a result the bottom cleats of truss T1 most likely broke away on impact with the ground whereas the threaded rods failed at the connections of T1 to T5 listed in Table 6.

Truss T9 also had its bottom cleat site welded similarly but its supporting post did not provide as much restraint.

The convention that appeared to have been used in the Drawings was that high strength bolts were designated by the use of “HS” as a modifier in the description (refer details S75-15, S75-23). Therefore elsewhere Property Class 4.6 rods and bolts would be assumed to have been specified as in the connection between the Community Courts Trusses and the spine truss T10 (refer detail S73-5, Figure 37).

The design shear capacity of a M20 Property Class 4.6 threaded rod in double shear was

fVfn = 44.6 x 2 = 90 kN

Failure Basis Evaluation of T1 to T9 Connections to Grid 4

In the Laboratory Examination the hardness of the threaded rods recovered from truss T1 (Figure 46 in Appendix C) were tested by MTL and tensile testing was performed by Uniservices. This found that the strength of the




threaded rods were compatible with those of stronger Property Class 6.8 rods.

The failure shear capacity of the M20 threaded rods using Property Class 6.8 properties in double shear was

Vfn = (44.6 x 2 x 600 / 400) / 0.8 = 167 kN

From Table 6 the top bolt connecting truss T1 and the bottom bolt connecting truss T5 to truss T10 was either very close to or just over the recommended design limit due to the moment fixity induced at the end of the T1 from the end cleats being welded preventing slippage of the bottom connection.

When the failure condition of G + 0.30 kPa was applied and the tested materials properties used for the threaded rods then it can be seen in Table 6 that the connections of the Community Courts trusses to Grid 4 were not the most critical components and did not initiate the collapse of the roof.




As-Built Design Basis

1.2G+1.2Su (Su=0.33 kPa)

Failure Basis

G+0.30 kPa

Truss Chord

Cleat

Axial

N*

(kN)

Shear

V*

(kN)

Resultant

Vr*

(kN)

Design

Capacity

Ratio

PC4.6

Resultant

Vrf*

(kN)

Failure

Capacity

Ratio

T1 TOP 138 0-79 138-159 1.53-

1.76*

110-127 0.66-

0.76*

T1 BOT 130 78-

157

NA N/A

(weld)

NA N/A

(weld)

T2 TOP 17 71 75 0.83 60 0.36

T2 BOT 0 71 73 0.81 58 0.44

T3 TOP 17 72 74 0.82 59 0.44

T3 BOT 0 72 72 0.80 58 0.43

T4 TOP 21 71 74 0.82 59 0.44

T4 BOT 0 71 71 0.79 57 0.43

T5 TOP 26 71 76 0.84 61 0.45

T5 BOT 0 71 71 0.79 57 0.43

T6 TOP 2 73 73 0.81 58 0.44

T6 BOT 0 73 73 0.81 58 0.44

T7 TOP 3 73 73 0.81 58 0.44

T7 BOT 0 73 73 0.81 58 0.44

T8 TOP 3 73 73 0.81 58 0.44

T8 BOT 0 73 73 0.81 58 0.44

T9 TOP 0 0-77 0-77 0-0.86 0-62 0-0.46

T9 BOT 11 77-

154

NA N/A

(weld)

NA N/A

(weld)




CONNECTIONS OF TRUSSES T10 AND T11 TO COLUMNS C13 TO C16

Design Basis Evaluation of T10 and T11 to Columns C13 to C16

The 1999 Review Report called for:

“Modified shoe details for columns C12, C13, C14 and C15” as shown on drawing 97139 (Figure 39) and

“A similar detail is to be used to release the bending on columns C13, C14 and C16. A minimum of four effective TCM connections to be provided on columns C13, C14 and C16.”

The detail shows the modified shoe bracket and welded mild steel packing plates for Trusses T12 to T15. Nothing was shown for the connection of T10 and T11 to columns C13 and C14.

The Site Examination found a similar arrangement however occurred there (Figure 40). Both ends of trusses T10 and T11 had welded packing plates approximately 40 mm thick at the bottom shoe bracket. The other mild steel packing plates up the end stanchion of the trusses were unwelded. This meant that only the bottom two TCM24 bolts would have been effective in transferring most of the vertical truss reactions into the column.

From Table 7 it can be seen that the bottom bolts in the connections of trusses T10 and T11 had as-built design capacity ratios greater than 1.0.

Failure Basis Evaluation of T10 and T11 to Columns C13 to C16

Hardness testing by MTL of a shoe bolt from C15 during the Laboratory Examination found the bolts to have average hardness values at the upper end of the range specified for Property Class 8.8 bolts, and at the upper end of Property Class 8.8 bolts indicating actual tensile strength of these bolts in the order of 1000 MPa (Hyland 2010).

From Table 7 it can be seen that the bolts in the connections of trusses T10 and T11 had failure capacity ratios less than 1.0.

The bolts connecting T11 to C13 and the TCMs connecting T10 to C14 fractured in tension. Whereas at the west end of the same trusses the bottom connections did not fail.




No lateral displacements

Westward lateral displacement

Discussion of T10 and T11 to Columns C13 to C16 Performance

The higher strength of the bolts meant these connections weren’t the initiator of collapse under the failure condition of G + 0.30 kPa.

The fact the bolted connections at C13 and C14 fractured in tension and those at the west end of the spine trusses connecting the trusses to C15 and C16 didn’t fail, can be explained by the superposition of gravity and drift induced bending moments on the frame shown in Figure 38.

The dead load and snow load induced actions on the spine trusses developed closing moments at the supporting column knees. The westward displacement imposed on the spine trusses and columns frame by the failing Community Courts trusses acted to reverse the closing moments at the east end and increase them at the west end. This led eventually to opening moments developing and tensile actions in the bottom chord shoe bolts at the connection to columns C13 and C14. These increased until bolt and TCM fracture occurred.

At the west end the westward displacement of the spine trusses increased the closing moment at the knee, so not adding any tensile actions to the connecting shoe bolts and TCMs at that end.

The bolts into the inserts at columns C15 and C16 did not fail in the collapse as the trusses were still connected to those columns after the collapse (refer Figure 42). Bolts from the inserts of C15 were hardness tested and found to have properties at the upper end of PC8.8. The bottom bolts or threaded rods into the M24 inserts into column C14 were sufficiently strong to fracture the inserts in tension rather than the bolts, indicating they were also stronger than the minimum specified for such inserts. The inserts are typically proportioned to be stronger than the bolts they accommodate. The M24 bolts or threaded rods connecting T11 into column C13 were observed to have a cupped fracture surface also indicating tensile fracture in combination with some shear. These fractures can be seen in Figure 40. The best explanation being that the bottom bolts connecting the spine trusses T10 and T11 to columns C13 and C14 fractured as the spine trusses were pulled westward and tensile actions




developed at those locations. This would have occurred after collapse initiated in the Community Courts trusses.




As-Built Design Basis

1.2G+1.2Su (Su=0.33 kPa)

Failure Basis

G+0.30 kPa

Column Truss Shear Actions

V* (kN)

Design Shear

Capacity

fVfn (kN)

Design Capacity

Ratio

Shear Actions

V*(kN)

Failure Shear

Capacity

Vfn (kN)

Failure Capacity

Ratio

C13 T11 328 266

(2 bolts)

1.23 262 416

(2 bolts)

0.63

C14 T10 434 266

(2 bolts)

1.63 347 416

(2 bolts)

0.83

C15 T11 552 399

(3 bolts)

1.38 442 624

(3 bolts)

0.71

C16 T10 415 266

(2 bolts)

1.56 332 416

(2 bolts)

0.80




CONCRETE COLUMNS C13 TO C16

Design Basis Evaluation of Columns C13 to C16

Spine trusses T10 and T11 were bolted to the corners of the supporting columns C13 to C16. The columns therefore would have experienced biaxial bending under normal service loads, in addition to axial actions. The combined design and failure actions have been tabulated in Table 8.

Cracking of these columns during construction was noted in the 1999 Review Report.

An axial load and bending moment interaction chart was generated for the 400mm x 300mm concrete columns C13 to C16 using the Section Analysis module of computer program ETABS (Figure 41). This interaction chart is based on design properties of Grade 430 steel reinforcement, concrete with minimum 28 day cylinder strength of 30 MPa, and 40mm cover to longitudinal reinforcing steel.

Two interaction curves are shown on the chart. The solid curve is for axial actions with bending about the major axis of the column and the dotted interaction curve is for axial actions with bending about the minor axis.

The axial action and bending moment data points shown in Table 8 have been plotted on the interaction chart for the NZS 4203:1992 load combination 1.2G+1.2 Su (Su=0.33 kPa).

These data points have been plotted for bending in each column axis independently, and ignoring slenderness effects. Slenderness effects magnify the bending moment actions on columns.

The Concrete Structures Standard only required interaction checks to be satisfied in each axis.

The strong axis design shear capacity of the columns was calculated to be

fVb=164 kN where the column ties were spaced at 150 centres.




Column Design Basis

1.2G+1.2Su (Su = 0.33kPa )

Failure Basis

G+ 0.30 kPa

Axial Action N*(kN)

Major Bending Moment

Mx*(kNm)

Minor Bending Moment

My* (kNm)

Axial Action N*(kN)

Major Bending Moment

Mx*(kNm)

Minor Bending Moment

My* (kNm)

C13 348 136 15 278 109 12

C14 455 99 38 364 79 30

C15 724 138 47 553 110 38

C16 436 161 12 349 129 10

Column C15 has the highest axial action by some margin as it supports one end of spine truss T11 and one end of trusses T14 and T15 that support the Events Courts roof (Table 8). All the columns have design actions that fit within the design interaction curve plot (Figure 41).

The 1999 Review Report included the following comments about eccentricity of connections to columns and column slenderness C15:

“The connections of the trusses to the columns are inadequate at present to support the full design loads due to eccentricity of the loading on the connection brackets. In addition the eccentric load is causing the support columns to bend and in some columns cracking has occurred. The bending capacity of the columns is adequate to support the full loads however there are some concerns regarding the slenderness of the columns.”

The Concrete Structures Standard NZS 3101:1995 required bending moments on slender columns to be magnified depending on the level of their slenderness. This was done either by carrying out a second order analysis, which took into account the additional bending induced in the column by the axial compressive action on the deflected shape, or by an approximate procedure based on the moment magnifier concept.

To assess slenderness an effective length factor ‘k’ was determined and applied to the unrestrained length to determine the effective length, ie taking into account the ‘fixity’ at top and bottom of the unrestrained length.




For many concrete structures a conservative effective length factor of k = 1.0 is often used in accordance with the recommendations of clause 8.4.11.2 in NZS 3101:1995 which states “The effective length factor, k, of a column or pier braced against side sway shall be taken as 1.0, unless analysis shows that a lower value may be used.”

An appropriate method of further analysis is the G-Factor Method (Bridge and Fraser 1987) which is the basis of the method described in the Appendices of the Steel Structures Standard NZS 3404:1997 (SNZ 1997).

Calculations by the authors using this method show that no moment magnifier effect was necessary for columns C13, C14 and C16.

For column C15 the effectiveness of the Control Room floor in providing lateral restraint affected the moment magnifier calculation for the design basis evaluation, but had no effect on the failure basis evaluation. For the failure basis evaluation there was no moment magnifier effect needed to be applied even if no lateral restraint was provided by the Control Room floor.

For the design basis evaluation, if no lateral restraint was considered to be contributed by the Control Room timber floor then the elastic critical buckling load Nc of column C15 was calculated to be 967kN for lu=10.3 m and k =0.7. With a design axial compression action of N*=724 kN a

moment magnifier m= 3.4 was required.

However if a lateral restraint capacity of 0.025N*=17 kN was provided by the Control Room floor then the unrestrained length of column C15 could be reduced.

It is not obvious from the Drawings whether it had been intended for the Control Room floor to provide this level of lateral restraint. The floor was shown on the Drawings as screwed and glued to floor joists which were bolted to steel 230 PFC sections that were bolted to the columns and adjacent walls each end with 2 –M16 anchors. In shear these would be expected to have a combined design capacity of at least 58 kN each, which was greater than the required lateral restraining action to the column. So a reasonable level of restraint could be expected from the Control Room floor.

For the design basis evaluation with lateral restraint provided by the Control Room timber floor the unrestrained length reduces to lu=6.6 m. The elastic critical buckling load Nc of column C15 was then calculated to be 1344 kN for lu=6.6 m and k =0.9. With a design axial compression action of N*=724

kN a moment magnifier m= 1.1 was required.

It was found that the reinforcement detailing in columns C13, C14, C15 and C16 did not strictly comply with clauses 8.4.7.2 (b) and (c) of NZS 3101:1995 as follows:

The maximum spacing of tie sets anywhere over the height of these columns should have been 100mm, and not 150mm or




300mm as shown on Eng Drawings S23 and S24 – to comply with NZS 3101:1995 clause 8.4.7.2(b); and

For columns of 400mm x 300mm cross section there should have been additional longitudinal bars placed at the mid length of each long side with additional cross ties around them - to comply with NZS 3101:1995 clause 8.4.7.2(c).

However this non-compliance is not considered by the authors to have initiated the collapse of columns C15 and C16. Columns C13 and C14 did not collapse.

The 3D structural analysis indicated that the shear demand on the columns under design snow loading was greatest in column C13 due to its attachment to the concrete wall panels. However this was only around 33% of the shear capacity of the column at the weakest locations where the shear reinforcing was R10 and 300 centres. So while the spacing of shear reinforcing did not satisfy the minimum requirements of the standard there was still sufficient capacity to sustain the design loadings in shear.

Failure Basis Evaluation for Concrete Columns C13 to C14

Columns C13 and C14 did not collapse.

Discussion of Concrete Columns C13 and C14 Performance

The eventual fracture of the bottom chord bolts connecting the spine trusses to these columns, (as the spine trusses were pulled westwards by the Community Courts roof (Figure 38)), occurred before the columns themselves were significantly damaged.

Failure Basis Evaluation for Concrete Columns C15 to C16

Columns C15 and C16 were loaded eccentrically by truss T10, T11 and truss T15. The columns appeared to have formed flexural hinges at the underside of trusses T10 and T11 and at the Control Room floor level. The top portion of these columns landed still connected to truss T10, T11 and truss T15 north and west of the base of the column. The bottom 2 to 3 metres of the column remained in place with longitudinal bars running between the two broken pieces (Figure 42).

For the failure basis evaluation, if no lateral restraint is considered to be contributed by the Control Room timber floor then the elastic critical buckling load Nc of column C15 was calculated to be 1344kN for lu=10.3 m and k =0.7. With a failure axial compression action of N*=603 kN a

moment magnifier m= 1.0 was required.

For the failure basis evaluation, with lateral restraint provided by the Control Room timber floor then the elastic critical buckling load Nc of column C15 was calculated to be 1526 kN for lu=6.6 m and k =0.9. With a failure axial




compression action of N*=603 kN a moment magnifier m= 1.0 was required.

Discussion of Concrete Columns C15 and C16 Performance

The fact that column C15 and C16 did not fail by weak axis bending at the Community Courts floor level (lu=10.3 m), but by strong axis bending at the Control Room floor (lu=6.6 m) shows that weak axis column buckling of C15 and C16 did not occur.

Discussion of Concrete Columns C13, C14, C15 and C16 Performance

It would have been desirable for columns C13 to C16 to be stiffer and stronger or to be braced to resist the westward drifts induced by the failing Community Courts trusses and prevent the collapse of the Events Courts roof area.

However it is beyond the current scope of design requirements for such collapse induced actions to be considered.



APPENDIX C: SUMMARY OF LABORATORY EXAMINATION RESULTS

INTRODUCTION

The following is a summary of the findings from the Laboratory Examination. A more detailed account is found in the Laboratory Examination report (Hyland 2010).

COMMUNITY COURTS ROOF TRUSS STEEL WALL THICKNESS

Portions of the Community Courts roof trusses T1 and T4 had wall thicknesses less than specified in the Drawings.

Some of the diagonal ties and vertical struts in truss T1 and T4 had wall thicknesses of 3.0 to 3.5 mm compared to 4.0 mm specified.

A portion of the bottom chord of truss T4 had a wall thickness of 5.2 mm compared to 6.0 mm specified.

WELD QUALITY

The welding was inspected by an experienced and CIBIP qualified welding inspector from MTL. The purpose of the inspection was to determine if the welds would comply with the Structural Steel Welding Standard AS/NZS 1554.1 Visual Examination criteria for Structural Purpose (SP) quality welds.

None of the welds observed were found to comply with the SP quality criteria in AS/NZS 1554.1.

COMPLETE PENETRATION WELD THICKNESS

The cross sectional area of the mid-span chord splices weld or Average Design Throat Thickness (“the DTT”) of the bottom chord of truss T1 and truss T9 were measured by HCL.

The weld thickness of the truss T1 bottom chord splice was 2.6 mm compared to the specified complete penetration weld DTT equal to the actual thickness of the truss chord of 5.9 mm.


APPENDIX C: SUMMARY OF LABORATORY EXAMINATION RESULTS continued


The DTT weld thickness of the truss T9 bottom chord splice (Figure 43) was 3.1 mm compared to the specified complete penetration weld DTT equal to the actual thickness of the truss chord of 9.4 mm.

The truss T1 bottom chord support cleat weld to truss T10 (Figure 44) was examined using metallographic techniques by Uniservices at UoA. The cleat had been site welded and the weld had fractured during the collapse. It had fractured in a brittle manner over a small portion of the failure surface and in a ductile manner elsewhere. The weld was undersized with thickness of 5.0 mm compared to that required being the 12.0 mm cleat thickness.

Flat fracture DTT min = 5.9 mm

Flat fracture DTT min= 4.1 mm

Flat fracture DTT min = 3.3 mm




WELD METAL PROPERTIES

The weld metal at the truss T1 and T9 bottom chord midspan splices was found from hardness testing by MTL to be stronger than the chord parent metal and the backing plug plate. This is consistent with the requirements of the Steel Structures Standard NZS 3404:1997.

SPINE TRUSS TO COLUMN BOLT PROPERTIES

The bolts connecting truss T11 to concrete column C15 (Figure 45) were found from hardness testing by MTL to conform to that required for Property Class 8.8 bolts as specified in the Drawings. Their hardness measurements were at the high end of properties for Property Class 8.8 bolts according to AS 4291.1-2000 (SAA 2000). They had properties consistent with the higher strength Property Class 10.9.




COMMUNITY TRUSS TO SPINE TRUSS THREADED ROD PROPERTIES

The bolts connecting truss T1 top and bottom chords to spine truss T10 (Figure 46) were found from hardness testing by MTL to conform to that required for Property Class 6.8 bolts rather than Property Class 4.6 specified. This meant they had greater capacity than was specified.

A tensile test of the rod confirmed it to have a 0.2% proof stress greater than the minimum required for Property Class 6.8 threaded rods.

COMMUNITY TRUSSES TOP AND BOTTOM CHORD TENSILE PROPERTIES

Tensile testing of steel from the 125 x 6 mm SHS top and bottom chords of Community Courts Trusses 1 and 4 at the Centre for Advanced Composite Materials by Uniservices (“the CACM”) found the chords to have tensile properties better than the minimum required for Grade C350 steel.

Tensile testing of steel from the 125 x 9 mm SHS top and bottom chords of Community Courts truss T9 found the chords to have tensile properties better than the minimum required for Grade C350 steel.

Tensile testing of steel from the 10 mm strengthening plates from the bottom chord of Community Courts truss T9 found the plates to have tensile properties consistent with that specified and required for Grade 300 flat bar.




PAINT CRACKING STRAIN

The strain at which the alkyd paint on the Community Courts truss chords became visually obvious was determined by testing at CACM by Uniservices to be on average 5.1% (Figure 47).




REINFORCING STEEL PROPERTIES

R10 column ties were found from testing at CACM to have yield stress, tensile strength and elongation properties in conformance with the requirements of Grade 300 reinforcing steel standard of the time NZS 3402:1989 (Figure 47).

20 mm deformed bars were found by full section testing at UoA by Uniservices to have stress, tensile strength and elongation properties in conformance with the requirements of Grade 430 reinforcing steel according to the standard of the time NZS 3402:1989 (Figure 48).




TRUSS TOP CHORD STRUT COMPRESSION BEHAVIOUR

Three truss chord specimens were prepared using salvaged chord remnants, with compression splices fit-ups, and strengthening side plate arrangements typical of those found in the Community Courts trusses. The top surface of the truss chords was left unwelded and proud of the splice plate, leaving three edges of the square hollow section in contact with the splice plate.

The specimens were tested by HCL in conjunction with Uniservices in compression, to failure at UoA (Figure 49).

The failures observed were similar to those observed in the collapsed roof structure in which the top surface of the square hollow section was cut into by the splice plate.

The average compression capacity was found to be 962 kN. This is well below the calculated nominal capacity of 1857 kN expected using the




average tested properties of the steel, if the chord compression splice had full contact on all four edges and strengthening side plates were installed as per the Drawings.

COLUMN CONCRETE COMPRESSIVE STRENGTH AND MODULUS

The mean concrete compressive strength of cores extracted from a remnant of concrete column C15, tested by Uniservices at UoA was 36.3 MPa. This is greater than the minimum specified 28 day strength of 30 MPa specified (Figure 50).

The 0.75f’cmax secant modulus was 32.2 GPa.

These values were used in assessing the buckling properties of the spine truss columns.

The concrete consisted of uncrushed river gravel aggregates.

0

5

10

15

20

25

30

35

40

0.000 0.001 0.002 0.003 0.004 0.005 0.006 0.007 0.008

Stre

ss M

Pa

Compressive Strain

Cored Cylinder Strength Column C-15

Core B

Core A



C11

C10

C9

C8

C7

C6

C5

C4

C3

APPENDIX D: A3 SIZE DIAGRAMS

3D MODEL OF ROOF STRUCTURE


APPENDIX D: A3 SIZE DIAGRAMS continued


TRUSSES 1 TO 9 MODIFICATION DRAWING


APPENDIX D: A3 SIZE DIAGRAMS continued


COLLAPSE SEQUENCE DIAGRAM

Purlins connected to T4 broke away from T3 from Grid 5 to midspan of T3

Top chord compression failure of T4 and T5 at quarter point splice That

had no end bearing plate

Spine trusses T10 and T11 bottom chord Connection 2-TCM 24 to Columns C13

and C14 fracture in combined tension and

shear

1500 mm vertical displacement of T1 after top chord compression failure at

splices of T1 to T5

Spine trusses T10

and T11 displaced

990 mm westward after collapse

T1 to T5 broke free from T10 upon impact with floor, deforming tie vertical struts in T10

Flexural compression hinging of columns C15 and C16 at underside of

T10 and T11 and at Control Room floor

Trusses T4 and T5 were pulled westwards off columns at Grid 5

Roof purlins and diagonal roof plane

bracing pulled collapsing Trusses T1

to T5 and roof westwards

T9 was pulled eastwards off column

support on Grid 5

T9 collapsed as bottom chord strengthening plates yield and

welds fracture

Likely tension failure of top chord splice from uplift of T9 due to

internal bursting pressure from collapse of T1 to T5, T10 and T11

C11

C10

C9

C8

C7

C6

C5

C4

C3

Glossary for terms used in the Stadium Southland Roof Collapse report Bending moment A measure of the amount of bending in a structural member Camber An allowance for future downwards movement, or deflection,

in a member Chord The top or bottom member of a framed beam or truss Cleat A piece of metal used to form a ledge or connection Collapse limit state The point at which a structure is expected to collapse Column head The top of a column Complete penetration butt weld

A butt weld in which fusion exists between the weld and the parent metal throughout the complete depth of the joint

Demand-to-capacity Ratio of the estimated force to the ability to resist a force DTT, design throat thickness

The ‘design throat thickness’ is the effective size of a full penetration butt weld, and is essentially the plate/pipe thickness of the thinner component

Fly brace Brace between a bottom chord and a supporting column kPa, kilopascal A unit of pressure or stress measuring force per unit area,

that has replaced the Imperial measurement system of pounds per square inch (psi)

Probability of exceedance

Estimated chance of a particular value being exceeded over a defined time period

Purlin, roof purlin Horizontal members laid to span across rafters or trusses, and to which the roof cladding is attached

Return period The average recurrence interval Shim Thin metal slip used to fill a space between structural

members – a very slim ‘packing plate’ SLS, Serviceability limit state

This state at which a building becomes unfit for its intended use through deformation, vibratory response, degradation or other physical aspects.

Sub-alpine Regions where the maximum snow load is usually due to a single downfall

Tributary area The area of building surface assumed to be supported by a structural element

Truss A frame to carry a roof or other load built wholly from members in tension and compression

ULS, Ultimate limit state

This condition is reached when the building ruptures, becomes unstable or loses equilibrium