th chapt 8 final

8-1

CHAPTER 8

SHAKING TABLE STUDY OF SINGLE-STORY DRY-STACK MASONRY

HOUSE

8.1 Introduction

Hydraform system now is widely used in Africa, Asia and South America. A

significant part of these continents lie in earthquake zones. Earthquake activity in

the eastern and southern Africa region is characterised by the occurrence of

destructive earthquakes, which are controlled by the well-known regional tectonic

feature, the East Africa rift system. In eastern Africa Uganda and Tanzania are

transversed by the East African rift system. Figure 8.1 shows the distribution of

earthquake epicenters in Africa region for the period 627 to 1994

[Turyomurugyendo (1996)]. The zones of active seismicity in southern Africa are

in Mozambique, Zimbabwe and northern Botswana. These activities are believed

to be an extension of the East African rift system.

The seismicity of areas located in the interiors of the major tectonic plates is low

but it is difficult to correlate with known tectonic characteristics

(Uzoegbo et al, 2000). The southern Africa region is located in an intra-plate area

but it is associated with a rather complex seismic characteristics. From Figure 8.1,

it is clearly seen that certain zones in the region have earthquake magnitude of

over 7, which gives a corresponding value of over 0.3 g ground acceleration.

However, most seismic events in the region measure below 5 on the Richter scale

and originate in the mining areas. Mine tremors occur at relatively shallow

epicenters, typically at depths of 2-4 km. This means that the spread is over a

small area. Moreover the attenuation of mine tremors is rapid. It is known that up

to 40 or more tremors are recorded monthly in South Africa. The recordings are

predominantly in the places surrounding the gold mining areas like the Transvaal

and Orange Free State. Many events are also recorded around the Carleton and

Klerksdorp areas annually. The tremors in the region have a characteristic of high

peak horizontal accelerations and velocities measuring up to 0.45 g and 67 mm/s

8-2

respectively. In Carletonville (1986) a peak acceleration of 0.45 g was measured.

This is even higher than the peak acceleration of 0.36 g obtained from the well-

known El Centro earthquake of 1940 recorded in USA. The difference between

the two seismic events is that while the El Centro Earthquake caused massive

damages to structures, the Carletonville tremor only caused minor damages and

cracking to structures located in the vicinity of the epicenter. Looking at other

characteristics of mine-induced tremors, they have high frequencies in the range

of 10 Hz to 50 Hz. This implies that mine tremors are not likely to produce

structural response from structures with low natural frequencies such as high-rise

buildings. A resonance, usually a response of large cyclic displacement amplitude,

is likely to occur when the excitation frequency is close to a natural frequency of

the structure. High-rise buildings do not respond to mine tremors because they

have low natural frequencies, typically less than 2 Hz. This means that single

story residential buildings are expected to respond to this type of excitation as

they fall in the high frequency range and hence the purpose for conducting the

shaking table study for the Hydraform system. Earthquakes of note, which

occurred in the 20th century in South Africa, are shown in Table 8.1.

Source NEIC-USA, 627-1994

Figure 8.1.Eathquakes in Africa continent

8-3

Table 8.1. Moderate earthquake occurred in 20th Century in South Africa Date Time Region Intensity/

magnitude

31/12/1932 08h32 Off Cape St.Lucia VIII/6.0-6.5

29/09/1969 22h03 Ceres/Tulbagh/Wolseley VIII-IX/6.3

08/12/1969 10h38 Welkom VII/5.5

07/03/1992 02h43 Carletonville VII/4.7

Source: Turyomurugyendo (1996)].

In the absence of standards for dry stacking, the experience from conventional

masonry will provide some guidelines in the development of the dry-stack seismic

system. Many researchers have reported the seismic performance of unreinforced

conventional masonry (Wesley, et al, 1986; Miha, et al, 1994; Petersen, at al,

2002; Pristley, at al, 1985; Minke, 2001; Martini, 1996; Jocelyn, 2001; Bolton,

1978).

The study was conducted to find if the normal Hydraform blocks and the normal

Hydraform construction system could resists the level of tremors in South Africa.

If so, is it adequate for resistance of tectonic activities, can the system be

improved for high intensity seismic zones. A full-scale single story dry-stack

interlocking block masonry building system was designed and tested on a shaking

table.

The test structure was subjected to a series of simulated earthquake ground

motions ranging from minor to severe. A conventional masonry test structure of

similar parameters was also constructed and tested for comparison. The tested

houses were monitored for peak acceleration [g], peak displacement [mm] and

mode of failure. Experimental data were captured and processed by computer

using continuous monitoring Software TLC CMSW 32 . The study was

conducted at SPOORNET laboratory in Johannesburg.

8-4

8.2 Object and Scope

The main objectives are to investigate the suitability of the Hydraform system for

the resistance of the level of seismic activities experienced in certain areas in

South Africa. The system would then be modified for design and construction in

more earthquake intense regions such as East and North Africa. The study

investigated the general structural behaviour of a single story unreinforced dry-

stack masonry house under seismic condition, the major structural weak points

were identified and strengthening alternative/methods were recommended.

8.3 Description of the Test Structures

8.3.1 House 1 Plain Dry-Stack Masonry

8.3.1.1 Wall construction details

House 1 was unreinforced dry-stack masonry (UDRM), constructed using

interlocking blocks without mortar or reinforcements except the starter course and

the lintel courses. The first test structure was viewed as an exploratory test for

providing the basic guideline in the identification of the weakest sections of

unreinforced dry-stack masonry under seismic load. The performance and mode

of failure of the structure forms basis for future modifications and strengthening

of the system. The test structures consisted of four walls, two with equal window

openings and two with equal door openings as shown in Figure 8.2. The

parameters of the structure are shown in Table 8.2.

a) East and North views b) West and North views

Figure 8.2: House 1 constructed on the shaking table

8-5

Table 8.2: House 1 parameters

External dimensions Doors Windows Remarks

Height

m

Width

m

Length

m

Height

m

Width

m

Height

m

Width

m

2.76 4.0 4.0 2.07 0.813 0.949 1022

Plain dry-stack,

lintel courses

in mortar

reinforced with

brick force

Single conduit seismic blocks (Figure 8.3a) and ordinary corner blocks (Fig.8.3b)

were used. The block compressive strength was 9 MPa. The compressive strength

was obtained by cutting 100 mm cubes from randomly sampled blocks and tested

for compression.

a) single conduit block b) ordinary corner block

Figure 8.3: Blocks for House 1

The Shaking table surface which forms part of the foundation of the test structure

is rigid therefore it was assumed that the tests would simulate only a very rigid

hard soil condition in the prototype. The foundation set-up (Fig. 8.5) was designed

to allow rocking and uplift but sliding was totally constrained. The starter course

was constructed on the Shaking Table in the 220 mm channel; the blocks were

laid in ordinary sand cement mortar class II and left for 3 days without load. The

channel was included in the table design to prevent sliding of the base of the

specimen during the test. Mid courses were dry-stacked, including the corners up

to the lintel level. The corners were plastered as shown in Figure 8.2. The top four

lintel courses were bonded using ordinary sand cement mortar class II, reinforced

with brick force and then plastered on both. Brick force by definition (Crofts,

8-6

2000) is a light welded mesh comprising of hard drawn wires of diameter not less

than 2.8 mm and not exceeding 3.55 mm held apart either by perpendicular

(ladder) or by diagonal (truss type) cross wires. In this investigation the diameter

of the wires was 2.8 mm. Figure 8.4 shows a roll of brick force. The blocks were

dry-stack in a stretcher bond.

Figure 8.4 Brick force

Ordinary steel window and door frames were used and were fixed in position in a

conventional manner followed by pre-cast concrete lintel above the windows. For

the purpose of accurate monitoring of the development of the initial failure of the

dry-stack walls, the structure was plastered inside using very thin (4-5 mm), weak

mix of sand: lime and painted white using water paint.

Figure 8.5: Starter course set-up on the Table

8-7

8.3.1.2 Roof structure and dead load

Lightweight steel trusses ( Z section) and purlins (channel type) were used. The

roof structure was assembled on the floor (Fig. 8.6(a)) and then lifted to position

by means of a crane. The 4 equally spaced trusses were tied down to the wall plate

by galvanised steel wires on top of the in-plane walls. 16 Concrete beams fixed

parallel with the purlins were added to the roof to simulate uniformly distributed

roof load as shown in Figure 8.6(b). Each beam was 4.5 m long of 16 kg/m self

weight. Lateral braces (3/4 flat steel bars) were provided between the trusses for

roof lateral stability.

In normal construction roof load varies with the square of the plan dimensions,

but perimeter length depend directly on these dimensions, so the average load per

unit wall length of a normal house varies linearly with the plan dimensions. The

test models were one-roomed structures. Unless the adjustments were made, the

roof load per unit length of the wall for the structure would have been very small.

So it was decided to attach concrete beams to the roof of the models in order to

achieve a design roof load of about 640 N/m², a minimum requirement according

to Code BS 6399 which states that minimum imposed roof load on a roof with no

access (where no access is provided to the roof (other than that necessary for

cleaning and maintenance) the minimum load is 0.6 kN/ m² for a roof slope 30

degree or less or a concentrated load of 0.9 kN.

8-8

a) assembling of the roof structure on the floor

b) additional dead load from concrete beams

Figure 8.6: Roof construction and dead load attachment

The structure was tested at the age of 3 month from the day the top courses were

laid

8.3.2 House 2 - Standard Conventional masonry House for comparisons

8.3.2.1 Wall construction details

House 2, a standard conventional masonry was constructed for comparison

purpose. The house was of similar parameters as House 1 and was tested in a

similar manner. Figure 8.7 shows, the North-West views of House 2 before

testing.

8-9

Figure 8.7: House 2 before test

Ordinary standard bricks of 6MPa compressive strength were used. The

dimensions of the bricks were 220 mm (L) x 100 mm (W) x 70 mm (T). The units

were laid in English bond to form one brick wall. Figure 8.8 shows the type of

bricks used.

Figure 8.8: conventional bricks

The foundation set up of was similar to House 1.The starter course was

constructed in a similar manner as in House1 (Fig.8.9) and the bricks were laid in

English bond using ordinary sand cement mortar (class II) and cured for 3 days

without load. Mid courses were also laid in English bond using ordinary sand

cement mortar class II and reinforced with brick force after every 3rd course. The

top four lintel courses were also reinforced with brick force similar to House 1.

Similar door and window frames were used followed by precast concrete lintels

above the windows. The structure was plastered internally and externally in

conventional manner using ordinary sand cement mortar class II, (25 mm thick)

8-10

and painted white to allow good monitoring of the initial cracks. See Figure 8.10

below. The structure was tested at the age of about 3 months.

Figure 8.9 Starter course set-up

a) plastered b) painted

Figure 8.10: wall surface treatments

8.3.2.2 Roof structure

Light weight steel roof structure of similar materials as in House 1 was used, with

additional lateral bracing between the trusses to enhance its capacity against

lateral movements. Figure 8.11 shows the assembling of the roof structure on the

floor in the laboratory before lifted to position using a crane.

8-11

a) assembling of the roof structure b) lateral brace between the trusses

Figure 8.11: Fabrication of the roof structure

8.4 Test Facility

8.4.1 Earthquake Simulator

8.4.1.1 Introduction

Because of high cost of a purchasing a shaking table following the quotations

from different supplies from different countries, and limited research funds

available, it was decided to design and fabricate a shaking table locally. The

shaking table was jointly designed by the WITS University and MEGMET (Pty)

Ltd. Conceptual design of the Shaking Table was based on the experiments

carried in phase one of this project. The table was designed to accommodate

similar size of test models as in Phase I; full-scale, one roomed single story

structure. A steel table was fabricated, mounted on recirculating roller type linear

bearings giving single degree of freedom (SDOF), providing motion in one

horizontal direction.

The table was installed at Spoornet laboratory owned by South Africa railways

Corporation at Jeppestown Johannesburg, where better facilities were readily

available. The laboratory is equipped with 10 Tonnes overhead crane, powerful

8-12

MTS Hydraulic system, modern data acquisition and processing system. The

laboratory building also has adequate headroom to accommodate a full size

single-story house on a Shaking Table.

8.4.1.2 Shaking Table specifications

Size: 4 m x 4 m

Maximum payload: 20 tonnes

Maximum height of payload centre of gravity: 1.65 m

Maximum horizontal acceleration (20 tonne payload): 1.0g

Maximum horizontal displacement: ±75 mm

Operational frequency range: 0-10 Hz

Table mass: 8 Tons

8.4.1.3 Design of the Table

Different configuration of the platform structure were modelled by means of the

finite element technique and investigated for natural frequency of vibration. The

platform structure was modelled by means of shell elements; whilst the specimen

was modelled rigidly as a point mass element. The point mass was modelled to be

rigidly connected to four points on the table surface, in between the four linear

bearing mounting positions. This assumption allows for maximum flexibility

between the connection points and the bearings and is thus conservative. A

symmetric half of the structure was modelled.

The liner bearings were modelled (by means of constraints) to be free in the

direction of movement of the table and rigidly fixed in the two perpendicular

directions. The first natural frequency was calculated by means of linear vibration

analysis (eigen value analysis).

The analysis shows that the most stiffness /mass efficient structure is achieved by

means of a solid plate top, supported by series of criss-cross T-sections, as

8-13

depicted in Figure 8.12. The results of analyses of different configurations along

this layout are shown in Table 8.3 below.

Table 8.3 Analysis results

Depth of

structure

(mm)

Web

thickness

(mm)

Top plate

and flange

thickness

(mm)

Mass of

specimen

(kg)

Height of

gravity centre

of specimen

above table

surface

(m)

First

natural

frequency

of loaded

table

(Hz)

First natural

frequency of

unloaded

table

(Hz)

Table

mass

(kg)

300 16 16 20000 1.3 19.4 49.0 5540

500 16 16 20000 1.65 24.0 67.9 6910

500 16 25 20000 1.65 27.8 68.4 8816

500 16 16

(covers

entire area)

20000 1.65 25.6 70.4 7548

ANSYS 5.3JAN 23 200216:17:41PLOT NO. 2ELEMENTSREAL NUM

1

11

XY

Z

XV =.934566YV =-1.119ZV =.934566DIST=2.412XF =.998997YF =.575ZF =2A-ZS=-.854E-06Z-BUFFEREDGE

Figure 8.12 Table construction

In all cases the mode shape, associated with the first natural frequency, was on

overturning motion of the specimen as shown in Figure 8.13.

8-14

ANSYS 5.3JAN 23 200216:16:49PLOT NO. 1DISPLACEMENTSTEP=1SUB =1FREQ=24.02RSYS=0DMX =.009849

1

11

XY

Z

DSCA=20.327XV =1YV =-1ZV =1DIST=2.393XF =.984179YF =.482655ZF =2.037Z-BUFFEREDGE

Figure. 8.13 Vibration mode shape

The second configuration in the table above (16 mm top plate, 16 mm flanges,

16 mm webs, 500 mm deep) was analysed for strength under application of full

weight of the table plus specimen, plus 1.0g horizontal acceleration. The results

are shown in Figure 8.14, as a contour plot of the von Mises equivalent stress.

Results show that the maximum stress is 60 MPa. Generally the stress is below 20

MPa in magnitude (thus 40 MPa range). With stress ranges not exceeding 40 MPa

good fatigue life will be achieved. Figure 8.15 shows the Shaking Table response

spectra.

8-15

ANSYS 5.3JAN 23 200216:22:11PLOT NO. 3NODAL SOLUTIONSTEP=1SUB =1TIME=1SEQV (AVG)TOPDMX =.553E-03SMN =51326SMX =.603E+08

1

MN

MX11

X

Y

Z

51326.675E+07.134E+08.201E+08.268E+08.335E+08.402E+08.469E+08.536E+08.603E+08

Figure 8.14 Stress distribution

Figure 8.15 Table response spectra

8-16

8.4.2 Data acquisition

The purpose of instrumentation was to obtain and record data from the hydraulic

and electronic equipment used in the experimental work. A computer installed

with monitoring system CMSW32 was used. CMSW32 is a powerful, user-

friendly continuous monitoring system for data acquisition and processing using

TLC software. The software is compatible to Microsoft Windows. Data is stored

in a database structure, which facilitates analysis and distribution. The data

acquisition system is capable of sampling up to 200 samples per sec per channel.

Acceleration was monitored using accelerometers and out of plane displacements

was monitored by Linear Velocity Displacement Transducers (LVDTs) all

connected to the data acquisition system shown in Figure 8.16. The filter

amplifiers were used as calibration tools for the signals from the displacement

measurement devices.

Figure 8.16 Spoornet data acquisition system

To prevent damage from falling debris the accelerometers were mounted in a

protection boxes fixed on the wall. The instrument locations are shown in Figure

8.17. Each location was monitored for accelerations and displacements except

location No 5, 13, and 14 which were monitored for accelerations only because of

shortage of instruments. The walls were monitored by ten LVDTs (D1 to D10),

8-17

and fourteen accelerometers (N1 to N14) but because of shortage N11 was moved

to the roof. The instrumentation frame was monitored by two accelerometers (N15

& N16). The table was monitored by one accelerometer (Atable). The location of

the monitoring positions for the two test houses remained the same. The walls

were monitored at the top; at mid height and at the bottom. The mid height

monitoring points were located either 500 mm from the nearest corner or 500 mm

from the nearest opening. The bottom monitoring points were located on the out-

of-plane walls at the 3rd course from the Table floor. The roof was monitored with

only one accelerometer, placed at the ridge. The Table accelerations were

monitored by accelerometer fixed at base of the table near the actuator

connections, while the Table displacements were monitored via inbuilt LVDT in

the actuator.

Figure 8.17: Monitoring Positions of the Test structures

N12

N13N14

1413

D8

12

911

10 8

7 6

5

4

3

21

NORTHWall type B (W3)

Wal

l typ

e B (W

4)

Wall type A (W1)

Wal

l typ

e A

( W

2)

of table motion

Direction

INSTRUMENTATION - INSIDE THE HOUSE

Ngowi WITS 2004

WEST

EAST

SOUTH

N4

N5

D4

N3

D10

N9D9

N10

N7

N6

D6

D7

N8

N1

N2D2

D1

D12

Atable

D3

RC beams supported on the purlins equally distributed

4 Light weight steel trusses equally spaced

Roof accelerometer at the ridge [N11]

N1-N16 - AccelerometersAtable - Table accelerometerD1-D10 - LVDTs

8-18

8.4.3 Reference frame

Displacements of the test structure relative to the shaking table movements were

measured from the rigid rectangular reference steel frame or instrumentation

frame, located inside the test structure, 100 mm from wall surface. The frame was

well braced and very stiff with designed natural frequency of 19.58 Hz for the

first mode. The maximum operating frequency of the Shaking Table is about

10Hz. It is important that the natural frequency of the table be not close to avoid

resonance. The frame was analysed using PROKON software (Fig.8.18a). LVDTs

were mounted on the Instrumentation frame. Figure 8.18b shows the Isometric

view of the stiff instrumentation frame, which was fixed on the table before the

construction of the test models. The presence of the instrumentation frame inside

the test specimen also provided protection in case of the structural failure of the

walls. To protect the actuators and other surrounding equipment in the laboratory

from falling debris, external protection frames were mounted around the test

models prior testing.

a) First mode natural frequency

Natural frequency for Mode no 1: 19.58 HzMaximum modal displacement at Node 10 in X direction

18

16

17

Y16

17

18

Y

21

19

20

21

24X

22

23

24X

15

13

14

15

28

29293030

27

25

26

27

12

10

11

12

1Z1

2

3

Z

6

4

5

6

9

7

8

9

8-19

b) 3D-rigid steel frame (section 76x2.5)

Figure 8.18 Reference frame

18

21

24

17

20

23

16

19

22

15

12

1

14

11

2

13

1028

3

30

6

295

4

27

9

26

8

25

7

Y

XZ

8-20

Figure 8.19: House 1 and 2 Parameters

8-21

8.5 Test Results

8.5.1 Introduction

The shaking table investigation was conducted to provide a basis for evaluation of

the structural behaviour of dry-stack masonry houses in a seismic condition.

Because of practical limitations in the implementation of typical details in the test

structures, only limited number of parameters could be included in the

investigation. The variables considered in the programme were the following;

i. Plain dry-stack walls

ii. Same base fixity for all specimens

iii. Roof orientation (remained unchanged with In-plane walls-load

bearing, out-of-plane walls non-load bearing in all tests)

iv. Standard conventional masonry system for comparison

The test structures were subjected to sinusoidal excitation. The sine sweep motion

used was of constant amplitude acceleration but varying frequency. In all tests the

table motions were applied sequentially, starting with very low intensity and

increasing until significant damage was observed. Because of this procedure the

structures were progressively damaged when subjected to the more intense

motions. Undoubtedly, the results obtained with these test structures may be

considered to be conservative in this regard.

There was interaction of the variables and consequently the determination of the

direct effect of a specific variable is difficult. One must therefore generalise in

order to bring out the more prominent features of the response characteristics. The

following section report accelerations and displacements recorded at various

positions on the test structure, mode of failure of the specimens and natural

frequency. Discussions and recommendations are also given.

8-22

8.5.2 Response of House 1

8.5.2.1 Input data

The shaking table test was performed using sine wave signals excitation to drive

the Table. In shaking table experiment this method is also known as a stroke

control method. The input variables to the computer to simulate the base motions

were frequency and stroke as indicated in Table 8.4. The test structure was

subjected to constant base motions with accelerations ranging from 0.05g to 0.3g

at varying frequencies ranging from 1Hz to 10Hz. A total of 31 sine wave signals

were applied. A chronological listing of the 31 test runs conducted on the

specimen including the physical site observations are provided in Table 8.5.

Table 8.4: Sine wave signal input data

Table Stroke [mm] Frequency [Hz] 0.05

g 0.1g 0.2g 0.3g 0.4g 0.5g 0.6g 0.7g 0.8g 0.9g

1 12.67

25.33

50.49

75.73

100.97

126.22

151.46

176.70

201.95

227.19

2 3.17 6.33 12.62

18.93

25.24 31.55 37.86 44.18 50.49 56.80

3 1.41 2.81 5.61 8.41 11.22 14.02 16.83 19.63 22.44 25.24 4 0.79 1.58 3.16 4.73 6.31 7.89 9.47 11.04 12.62 14.20 5 0.51 1.01 2.02 3.03 4.04 5.05 6.06 7.07 8.08 9.09 6 0.35 0.70 1.40 2.10 2.80 3.51 4.21 4.91 5.61 6.31 7 0.26 0.52 1.03 1.55 2.06 2.58 3.09 3.61 4.12 4.64 8 0.20 0.40 0.79 1.18 1.58 1.97 2.37 2.76 3.16 3.55 9 0.16 0.31 0.62 0.93 1.25 1.56 1.87 2.18 2.49 2.80 10 0.13 0.25 0.50 0.76 1.01 1.26 1.51 1.77 2.02 2.27 11 0.10 0.21 0.42 0.63 0.83 1.04 1.25 1.46 1.67 1.88 12 0.09 0.18 0.35 0.53 0.70 0.88 1.05 1.23 1.40 1.58

8-23

Table 8.5: Test sequence House 1-UDRM

File Name Frequency [Hz]

Base motion [g]

Physical site observations (progressive damage)

1 2 3 4

Sine 01-005-Test 1 1 Sine 02-005-Test 2 2 First crack on the plaster (<1mm) below point 2 West

wall

Sine 03-005-Test 3 3 Sine 04-005-Test 4 4 Initial crack on the plaster (< 1mm) below widow sill

South wall and near point 5 North wall

Sine 05-005-Test 5 5 Sine 06-005-Test 6 6 Sine 07-005-Test 7 7 Sine 08-005-Test 8 8 Sine 09-005-Test 9 9 Sine 10-005-Test 10 10

0.05

Sine 10-01-Test 11 10 Sine 09-01-Test 12 9 Initial diagonal cracks on the plaster (1mm) at lintel

course East wall and vertical crack (1mm) near point 1 West wall

Sine 08-01-Test 13 8 Widening of the previous cracks at East and West wall to about 2 mm and 2 mm crack at lintel course North wall

Sine 07-01-Test 14 7 Falling of plaster near point 7 East wall

Sine 06-01-Test 15 6 Widening of the previous diagonal crack and falling plaster at lintel course East wall

Sine 05-01-Test 16 5 Initial vertical crack at the East/North corner junction from the top to mid height

Sine 04-01-Test 17 4 Sine 03-01-Test 18 3 -External diagonal crack at lintel level East wall

Sine 02-01-Test 19 2

0.1

- Very intense roof movement

Sine 10-02-Test 20 10 Previous diagonal crack at East wall extending from the top the widow sill level

Sine 09-02-Test 21 9 -Continue of falling plaster form lintel courses East -East / south wall, corner junction failure 2 courses from the top - West/ North corner junction failure from the top to the lintel level.

Sine 08-02-Test 22 8 Sine 07-02-Test 23 7 Sine 06-02-Test 24 6 Extension of the corner cracks on both sides of East

wall from the top to window sill level -sliding of the units at lintel courses (about 6 mm)

Sine 05-02-Test 25 5 Sine 04-02-Test 26 4 Falling of external plaster at lintel level West wall

Sliding of the lintel and the units at lintel level (gape about 15 mm) -Development of diagonal crack from the mid height at the corner to the bottom near door frame at West and North wall

Sine 03-02-Test 27 3 Rotation and sliding of the corner blocks at East wall


0.2

Bending of the door frame; opening of the vertical joints West wall (10 mm) along failure line

Sine 01-01-Test 29 1 0.1 Sine 01-02-Test 30 1 0.2 Total damage of west /east wall corners from top to

widow sill level; vertical joints opening 20 mm; Permanent outwards deflection of East/West wall (75 mm)

Sine 01-03-Test 31 1 0.3 Partial collapse of East / West walls

8-24

8.5.2.2 Response of the test structure.

The first series of tests (T1 - T10) was conducted using a constant base motion of

0.05 g with varying frequency range of 1Hz to 10Hz. Before testing the structure,

all the hairline cracks inside the structure were marked. The hairline cracking was

likely due to the shrinkage of the plaster during the curing period. The roof

structure was supported on in-plane walls (see Fig. 8.2 and 8.19), trusses oriented

perpendicular to the table motion. The out-of-plane walls were therefore non-load

bearing. The cracks on the plaster, caused by the base motions were initially

Figure 8.20: Initial failure West wall at 2Hz [0.05g]

a) South wall b) North wall

Figure 8.21 Initial failure at 4Hz [0.05g]

recorded at frequency of 2Hz and 4Hz on the mid sections of the West and North

walls. On the South wall the initial cracks were noted below the window sill.

Figure 8.20 and 8.21 shows these initial failures.

8-25

Peak displacement and acceleration recorded on the top of the out-of plane walls

and In-plane walls are presented in Table 8.6 and 8.7 below respectively. The

maximum displacement amplitudes were recorded at the top the walls at lower

frequencies. The displacements were of the order of 0.26 to 4.07 mm in out of

plane walls and 0.1 to 1.49 mm in In-plane walls. The recorded response

quantities for T2 and T4 are presented as time histories in Figure 8.42 and 43.

This should be studied in conjunction with Table 8.5.

The second series of tests (T11- T19) was conducted using a constant base motion

of 0.1g-frequency range 2Hz to 10Hz. During the tests, significant damage to the

structure such as cracks on the plaster at the lintel courses, initial development of

diagonal cracks on the East wall, falling of the plaster and initial corner failure of

East-North walls were observed. Figure 8.22- 8.25 shows these failures. The

maximum displacements at the top of the walls were of the order of 1.35 to

9.0 mm in the out-of-plane walls and 0.61 to 4.01mm in the In-plane walls as

shown in Table 8.6 and 8.7. Figure 8.44 shows the time histories of the structure

at Test 12.

a) East wall b) West wall

Figure 8.22: Failure pattern at 9Hz [0.1g]

8-26

Figure 8.23: North wall failure at 8Hz [0.1g] Figure 8.24: corner failure at 5Hz [0.1g]

Figure 8.25: East wall failure at 6Hz [0.1g]

The third series of tests (T20 - T28) was conducted using a constant base motion

of a 0.2g , frequency (2Hz -10Hz ). These series of tests caused many noticeable

damages to the structure such as widening and extension of the existing diagonal

failures at East and West walls, rotation and sliding of the units at lintel courses at

West wall likely due to hinging action at upper mid section of the out-of-plane

walls, initial failure of the corner connecting West and North walls (Fig. 8.26).

The hinging action of the out-of plane walls captured during the test is quite

visible (Fig.8.27). Test 28 at 2Hz was very destructive resulting into bending of

the door frame in West wall (Fig. 8.28). Test 30 at 1Hz caused very destructive

out-of plane displacement of the East and West walls with maximum deflection of

about 62 mm recorded at the top of West wall.

It was observed that the diagonal failure of the dry-stack masonry walls occurred

mainly by the opening of the perpend dry-stack joints and not through units

failure. Figure 8.29 shows the typical diagonal failure pattern before partial

8-27

collapse of out-of-plane walls. Figure 8.45 and 8.46 shows the time histories of

the structure during Test 21 and Test 28 respectively.

The fourth series of tests was conducted using 0.1g, 0.2g and 0.3g base motions at

1Hz. Test 31 caused almost total collapse of the out-of-plane walls as shown in

Figure 8.30. The Figure 8.31 and 8.32 shows the failure details of the out-of-plane

walls. East wall suffered severe damaged with about 40% collapse followed by

West wall with 8% collapse with out-of-plane deflection of about 100 mm. The

in-plane walls remained in position with only major cracks (opening of perpend

joints) at the lintel courses including rotation the corner blocks from the top to the

bottom courses. The maximum out-of-plane deflection of In-plane walls was

about 35 mm. Figure 8.47 shows the time histories of the test structure at T30.

Figure 8.26: Initial corner failure at 9Hz [0.2g]

Figure 8.27 East wall deflection at 1Hz [0.2g] Figure 8.28 Bending of the door frame 2Hz [0.2g]

8-28

a) door opening b) window opening

Figure 8.29 Typical diagonal failure of the walls

a) East wall b) West wall

Figure 8.30 collapse of out-of-plane walls at 1Hz [0.3g]

Figure 8.31 East wall failure details

Figure 8.32 West wall failure details

8-29

The dry stack masonry exhibited some flexibility and therefore exhibits some

vibratory response mechanism. The significant aspect of these mechanisms is that

the applied table motions were amplified by the structural response so that peak

acceleration recorded at the top of wall were much greater than input acceleration.

This phenomenon is well reflected in Table 8.6 and 8.7 columns 8, 9, and 13.

8.5.2.3 Performance of the roof structure

The trusses were tied on top of timber wall plate using 5 mm galvanised mild steel

wires enchored to the supporting walls; at 3rd or 4th course below the wall plate

(Fig.8.33). The trusses were therefore held in position but not restrained from

movement i.e. sliding. Additional dead load (concrete beams) added was

necessary to simulate the shear forces generated during earthquake in a normal

house of normal floor plan size. The additional beams provided a reasonable

approximation of the both dead load per unit length of load bearing wall and the

inertia load transferred to the in-plane walls as the plan size of the test structure

was limited by the size of the Shaking Table.

Figure 8.33 Roof wire - mid trusses

In all experiments the trusses were oriented perpendicular to the direction of

excitations; therefore the effect of the trusses parallel to the base motions cannot

be discussed in this investigation. The roof was monitored by only one

accelerometer placed at the ridge. There was no displacement gauges available,

therefore the roof displacements was not monitored. The roof connection used in

this investigation is the one normally used in non-seismic regions and where wind

8-30

load is considered to be normal for single story houses. Roof connection using

wires is considered to be technically and economically viable in low cost housing

projects; cheap and easy to install and structurally sound. In this investigation this

type of roof connection was assumed to simulate the worst scenario, and the

results obtained will in future be used in the development of appropriate roof

connection for seismic condition.

The roof was supported on the In-plane walls and the maximum accelerations

recorded at the top of the walls and at the roof ridge is shown in Table 8.7. The

results indicate that the roof recorded recoded higher accelerations compared to

the supporting walls. As the base motions intensifies and the structure continue to

weaken the roof accelerations amplitudes also increases as shown in Figure 8.34

below. There was an increase of acceleration amplitudes of between 42% to 70%.

Recorded max. accelerations at the roof and supporting walls

0.58

0.97 0.93

0.60

1.01

1.64

1.90

2.10

0.00

0.50

1.00

1.50

2.00

2.50

0.05g 0.1g 0.2g 0.3g

Base motion

max

. acc

eler

atio

ns [g

]

wall roof

Figure 8.34 Recorded max. accelerations at the roof and the supporting walls

The response of the roof against the supporting walls and the base during the

0.05g test runs is given in Figure 8.35 below. The time history for the roof

structure response is also shown in Figure 8.42 to 8.47.

8-31

Peak accelerations House 1 at 0.05g

0

0.2

0.4

0.6

0.8

1

1.2

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]

Noth wall South wall Table roof

Figure 8.35 Roof response against the base and the supporting walls

The roof structure was relatively flexible and bracing provided between the

trusses was not adequate. The roof failed by bending and twisting of the rafters

and the bottom cords of the trusses; buckling of the lateral braces and shear failure

of the connections where the screw threads were weakened due to fatigue as

shown in Figure 8.36 below.

a) Bending of the truss

8-32

b) Bending of rafters and twisting of the bottom chords

c) Typical Pull-through Failure at Screw Connection

Figure 8.36 Truss failure -House1

8-33

8.5.3 Natural frequency of the test structure - House 1

Normally in conventional masonry the magnitude of building response, that is the

acceleration, which it undergoes, depends primarily upon the frequencies of the

input ground motion and the building s natural frequency (Tena, 1992, Clough et

al, 1979). When these are near or equal to one another, the building s response

reaches peak level. In some circumstances, this dynamic amplification

effect

can increase the building acceleration to a value much greater than the

acceleration at the base of the building. This phenomenon was also observed in

dry-stack masonry tested as reflected in Table 8.6 and 8.7.

The relation between frequency f and period T is given by T = 1/ f. Fundamental

period T is an inherent property of a building, any alterations made to the building

will change its T. This phenomena was also observed in dry-stack masonry tests

as the structure continue to weaken due to structural damage the fundamental

frequency also changes and hence the period.

Based on the test results of the first test series at 0.05g given in Table 8.6 column

9 the maximum dynamic amplification at East wall House 1 was 5.11 times.

Figure 8.37 suggest that the natural frequency of House 1 at 0.05g was about 9Hz,

implying that its natural period T drywas 0.11 sec.

House 1 Plain dry-stack peak acceleration East wall at 0.05g

0.0

0.5

1.0

1.5

2.0

2.5

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]

Top middle near window

middle near corner bottom

a) out-of-plane wall

8-34

House 1 Plain dry-stack

peak acceleration North wall at 0.05g

0.0

0.5

1.0

1.5

2.0

2.5

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]

Top middle near corner middle near door

b) In-plane wall

Figure 8.37: Natural frequency House 1 at 0.05g

At the second series of tests of 0.1g base motion, the natural frequency of House 1

was about 8Hz, (Fig.8.38) implying that the fundamental period of the building

had increased to 0.12sec. The maximum dynamic amplifications recorded at out-

of-plane walls was 7.60 times, an increase of 48% when compared to

amplification in the first series of test at 0.05g.

House 1 peak acceleration West wall at 0.1g

0.0

0.5

1.0

1.5

2.0

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]

Top middle near door middle near corner

a) out-of-plane walls

8-35

House 1 North wall peak acceleration at 0.1g

0.0

0.5

1.0

1.5

2.0

0 1 2 3 4 5 6 7 8 9 10 11Frequency [Hz]

peak

acc

eler

atio

n [g

]

Top middle near corner middle nead door

b) In-plane walls


In the third series of tests at a constant base motion of 0.2g, the test structure

recorded almost random vibration at different levels (Fig. 8.39) likely due to the

progressive damages from the previous tests. From the test results the natural

frequency of House 1 at 0.2g could be estimated to be about 3Hz, implying that

the fundamental period was 0.33 seconds, three times the period of the structure at

0.05g the first test series. In the final experiments the maximum dynamic

amplifications was 8.37, which was recorded at West wall. This was an increase

of 63% compared to the first experiments before the test structure was damaged. .

House 1 peak acceleration East wall at 0.2g

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

1.6

1.8

2.0

0 1 2 3 4 5 6 7 8 9 10 11Frequency [Hz]

peak

acc

eler

atio

n [g

]

Top middle near window middle near corner bottom

a) out-of-plane walls

8-36

House 1 peak acceleration North wall at 0.2g

0.00.20.40.60.81.01.21.41.61.82.0

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]


b) In-plane walls


8.5.4 Discussion

House 1 was unreinforced dry-stack masonry (UDRM), and was considered to be

the weakest form of dry-stack masonry system. It was investigated first to provide

basis for further improvement and strengthening of the system. The test models

were subjected to sinusoidal excitations, which according to Crewe and Daniell

(2004) has some advantages. In a shaking table test, large random earthquake

motions tends to visually mask the smaller motions of the structure; However, the

mode of shape and structural amplification are readily apparent when a model is

subjected to sinusoidal excitation at its natural frequency.

Test results suggest that the major cause of failure of House 1 was inadequate

lateral resistance capacity against out-of-plane forces. The out-of-plane walls,

which were also non-load bearing, were the most vulnerable. The walls

experienced excessive out-of- plane displacements at the upper mid-section

(Fig.8.27), causing sliding and rotations of the interlocking units (Fig.8.40) and

ultimately corner failures (Fig.8.41), which lead to the partial collapse of the

walls. The test results also suggest that the progressive damage of the test model

reduces its stiffness and hence the natural frequency. At the point of failure during

the third series of tests the natural period of the structure was 0.33 seconds

compared to 0.11 seconds in the first series of the tests, a three times increase.

8-37

Based on the test results, the following are the major structural problems

identified and need to be addressed in order to improvement the structural

capacity of plain dry-stack masonry system under seismic condition: -

1. Rotation and sliding of corner blocks.

2. Excessive out-of-plane displacements of the walls.

3. Rotation and sliding of precast lintels.

4. Openings between the dry-stack perpend joints and sliding of the

interlocking walling units.

5. Bending of the door frames.

6. Weak connections between the In-plane and Out-of-plane walls

7. Falling debris.

8. Weak roof anchorage to the supporting walls.

9. Local buckling of the trusses and shear failure of the roof connections.

Figure 8.40. Sliding and opening of the perpend joints in the mid courses-House 1

8-38

Figure 8.41. Typical corner failure House 1 [0.3g]

8-41

Response House1 - UDRM

a) Table acceleration T2

N1 (top) N2 (mid) West wall N7 (top) N6 (mid) - East wall

b) Acceleration at the top and mid height of out-of-plane walls T2

8-42

N4 (top) N3 (mid) North wall N10 (top) N9 (mid) South wall

c) Acceleration at the top and mid height of In-plane walls T2

d) Roof acceleration T2

Figure 8.42 Measured response: House 1, test 2 [0.05g @2Hz]

8-43




8-44




Figure 8.43 Measured response: House 1, test 4 [0.05g @4Hz]

8-45




8-46




Figure 8.44 Measured response: House 1 test 12 [0.1g @ 9Hz]

8-47




8-48




Figure 8.45 Measured response: House 1, test 21 [0.2g@ 9Hz]

8-49




8-50




Figure 8.46 Measured response: House 1, test 28 [0.2g @ 2Hz]

8-51




8-52




Figure 8.47 Measured response: House 1, test 30 [0.2g @ 1Hz]

8-53

8.6 Response of House 2

8.6.1 In put data

The structure was tested in a similar manner to House 1, using sine wave signals

(Tab.8.4). The structure was subjected to constant base motions ranging from

0.05g to 0.8g at frequency range of 1Hz to 12Hz. A total of 65 sine wave signals

were applied. The chronological order of test sequence and site physical

observations are presented in Table 8.8 below.

8.6.2 Response of the test structure

The first series of tests (T1-T10) was conducted using a constant base motion of

0.05g with varying frequency of 1Hz to 10Hz. As in House1 the out-of-plane

walls (West and East) were non-load bearing and In-plane walls (North and

South) were supporting the roof structure and therefore load bearing. A total of

ten sine wave signals were applied. The recorded out-of-plane displacements

amplitudes of out-of-plane walls were of the order of 0.17 mm to 0.97 mm, while

in-plane walls were up to 0.72 mm. The maximum acceleration of out-of-plane

walls was 2.0g recorded at West wall and both In-plane walls recorded maximum

acceleration of 0.82g. The peak displacements amplitudes and acceleration at the

top of the out-of-plane and in-plane walls are given in Tables 8.9 and 8.10

respectively. The tables also show the roof peak accelerations recorded at the

ridge level. The maximum roof acceleration recorded during 0.05g base motions

was 0.41g. No noticeable structural damage to the test structure was recorded in

the first series of the tests.

8-54

Table 8.8 Test sequence House 2 conventional masonry

File name Freq. [Hz]

Base Motion

[g]

Physical site observations

Sine 01-005-Test 1 1 Sine 01-005-Test 1 1 Sine 02-005-Test 2 2 Sine 03-005-Test 3 3 Sine 04-005-Test 4 4 Sine 05-005-Test 5 5 Sine 06-005-Test 6 6 Sine 07-005-Test 7 7 Sine 08-005-Test 8 8 Sine 09-005-Test 9 9 Sine 10-005-Test 10 10

0.05g No damage

Sine 10-01-Test 11 10 Initial horizontal crack about (1mm) to the starter course -North wall

Sine 09-01-Test 12 9 Sine 08-01-Test 13 8 Sine 07-01-Test 14 7 Initial vertical cracks (about 1mm) East wall at lintel courses

Sine 06-01-Test 15 6 Sine 05-01-Test 16 5 Sine 04-01-Test 17 4 Sine 03-01-Test 18 3 Very strong horizontal movement of the roof structure in the direction

of the base motion.


0.1g

Sine 12-02-Test 20 12 Sine 11-02-Test 21 11 Sine 10-02-Test 22 10 Initial vertical cracks (about 1mm) North wall at lintel courses and

development of horizontal crack along the bed joints, of the starter course

Sine 09-02-Test 23 9 Rocking of the structure and extension of the previous cracks at the lintel level North wall

Sine 08-02-Test 24 8 Sine 07-02-Test 25 7 Sine 06-02-Test 26 6 Falling of the plaster from the starter course West wall

Sine 05-02-Test 27 5 Initial horizontal crack (2mm) along bed joints of the starter course East wall

Sine 04-02-Test 28 4 Sine 03-02-Test 29 3 Very strong horizontal movement of the roof structure in the direction

of the base motion.


0.2g

Sine 10-03-Test 31 10 Increase of the of the existing cracks at North wall at lintel courses to about 2mm

Sine 09-03-Test 32 9 Initial vertical cracks (about 2mm) at lintel courses West wall

Sine 08-03-Test 33 8 Falling plasters from starter course East wall and widening of the existing horizontal cracks

Sine 07-03-Test 34 7 Sine 06-03-Test 35 6 Rocking and lifting of the structure and further damage to the starter

courses North East walls

Sine 05-03-Test 36 5 Sine 04-03-Test 37 4 Sine 03-03-Test 38 3


0.3g

Complete isolation of the structure from the base due to horizontal cracks along starter course bed joints at West North - East walls;

Base uplift of up to about 6mm is observed during the rocking of the structure.

8-55

Table 8.8 continues

The second series of tests (T11-T19) was conducted using a constant base motion

of 0.1g with frequency range of 10Hz to 2Hz. At sine wave signal of 10 and 7Hz

initial failure was recorded at North and East walls respectively. The North wall

developed a horizontal crack (about 1 mm) wide at the starter course along the

bed joints as shown in Figure 8.48(a). At East wall the initial failure was 1mm

cracks on the plaster at the lintel courses above the window as shown Figure

8.48(b). A maximum acceleration of 2.74g was recorded at West wall with

File name Freq. [Hz]

Base Motion

[g]

Physical site observations

Sine 10-04-Test 40 10 Sine 09-04-Test 41 9 Widening of the existing vertical cracks at lintel courses of North wall

to about 6 mm

Sine 08-04-Test 42 8 North wall base up lift of about 8 mm is noted

Sine 07-04-Test 43 7 Sine 06-04-Test 44 6 Widening of the existing vertical cracks at lintel courses of West wall to

about 4 mm

Sine 05-04-Test 45 5 Sine 04-04-Test 46 4 North wall base up lift of about 10 mm is noted


0.4g

Diagonal failure of East wall from the top to the bottom (2mm to 6mm cracks

Sine 10-04-Test 48 10 Sine 09-04-Test 49 9 Sine 08-04-Test 50 8 Widening of the existing diagonal cracks East wall to about 8mm

Sine 07-04-Test 51 7 Sine 06-04-Test 52 6 Sine 05-04-Test 53 5 Widening of the existing diagonal cracks East wall to about 12 mm

Sine 04-04-Test 54 4 Sine 03-04-Test 55 3

0.5g

Sine 10-04-Test 56 10 Widening of the existing diagonal cracks East wall, 15 mm to 20mm

Sine 09-04-Test 57 9 Sine 08-04-Test 58 8 Sine 07-04-Test 59 7 Sine 06-04-Test 60 6 Widening of the existing cracks at lintel courses North wall -30 to 40

mm

Sine 05-04-Test 61 5 Collapse of lintel courses East wall above the window

Sine 04-04-Test 62 4 Window frame fallen out


0.6g

Sine 03-07-Test 64 3 0.7g Development of corner vertical cracks from the top to mid height North wall, bottom to mid height West wall and permanent base uplift of about 12 mm -North wall

Sine 04-08-Test 65 4 Further destruction of East wall

0.8g

8-56

corresponding out-of-plane displacement of 1.32 mm as shown in Table 8.9. The

maximum roof acceleration was 1.14g

(a) Initial failure North wall at 10Hz [0.1g]

(b) Initial failure East wall at 7Hz [0.1g]

Figure 8.48 Distribution of cracks: House 2 at 0.1g

The third series of tests (T20-T30) was conducted using a constant base motion of

0.2g, frequency (2Hz to 12Hz). This range of test signals caused significant

structural damage to the specimen. At 10Hz signals (T22), initial failure of the

North wall at lintel level was observed including further extension of the existing

horizontal crack at the starter course along the bed joints as shown in

Figure 8.49(a). Initial failure at West wall (Fig.8.49 (b)) was recorded at 6Hz

signals in T26. At 5Hz signals (T27), initial failure of the starter course at East

wall was observed (Fig.8.49 (c)). At T29 very strong lateral movement of the roof

structure was observed. The recorded responses for T22 are presented as time

histories in Figure 8.63. This should be studied in conjunction with Table 8.8

8-57

a) Failure North wall at 10Hz [0.2g] b) Initial failure West wall at 6Hz [0.2g]

c) Initial failure East wall at starter course 5Hz [0.2g]

Figure 8.49 Distribution of cracks on the walls: House 2 at 0.2g

The fourth series of tests (T31-T39) was conducted using 0.3g constant base

motions with frequency ranging from 2Hz to 10Hz. At 9Hz signals (T32), the

west wall developed initial vertical cracks at the lintel level (Fig. 8.50) Rocking of

the test structure was also observed, which resulted into development of

horizontal cracks at starter courses through out the West-North-East walls,

isolating the super structure from the foundation in T37. This was result of a shear

failure of the starter courses just above the steel channel, which restrain the base

from sliding. During the test runs an uplift of the test structure base, of up to 6

mm was observed in the In-plane-East wall. The time histories recorded at T32

and T37 are shown in Figure 8.64 and 8.65 respectively.

Figure 8.50 Initial failure at lintel courses West wall [0.3g @ 9Hz]

8-58

The fifth series of tests (T40-T47) was conducted using a constant base motion of

0.4g with frequency range of 10 to 3Hz. The test structure continued to rock with

base uplift of about 8 mm noticed at the base of North wall. The existing cracks

continued to widen, and at T47, the East wall developed a continuous diagonal

failure from the top of the wall to the bottom. The widths of the diagonal cracks

were from 2 to 5 mm at the top to 6 mm at the bottom courses as shown in Figure

8.51 below. The response of the test structure at T47 is shown as time histories in

Figure 8.66.

Figure 8.51 Diagonal failure East wall at 3Hz [0.4g]

The sixth series of tests (T48-T55) was conducted using a constant a base motion

of 0.5g with varying frequency range of 10Hz to 3Hz. The tests caused further

structural damage by extending and widening the cracks at East wall to about

12 mm wide as shown in Figure 8.52.

Figure 8.52 Diagonal failure at East wall [0.5g @ 5Hz]

8-59

The seventh series of tests (T56 - T63) was conducted using a constant base

motion of 0.6g at varying frequency range of 10Hz to 3Hz. The tests were very

destructive likely because the structure was already weakened by the previous

tests. At 6Hz test signals (T60), the existing cracks at North wall widened to

30-40 mm. At 5Hz test signals (T61), collapse of the lintel courses above the

window was observed (see Figure 8.53). At 4Hz test signals (T62), the widow

frame from East wall fell out (Fig.8.54). The last two final tests were conducted

using 0.7g @ 3Hz and 0.8g @ 4Hz causing further destruction to the structure. A

permanent base up lift of 12mm was recorded at the North wall base at the final

test. See also appendix E-2 for extra details of wall failure.

Figure 8.53 Collapse of the lintel courses East wall at 5Hz [0.6g]

Figure 8.54 window frame out of the structure

8-60

8.6.3 Performance of the roof structure

The trusses were tied to the supporting In-plane walls in a similar manner as in

House 1 using wires. Same additional dead load (concrete beams) were added to

the roof structure as in House1. The roof was monitored in a similar manner using

only one accelerometer positioned at the ridge level. The roof structure was well-

braced using lateral bracing as shown in Figure 8.11(b). The roof withstands base

motions intensity of up to 0.8g (about 7.7 Richter scale), with minimum structural

damage to the rafters at the supports. The rafters failed by local buckling of the

sections at the supports (Fig.8.55), the point of contact with wall plate. The rest of

the roof structure remained without any significant structural damages to the

members and connections as shown in Figure E-5 in Appendix E.

Figure 8.55 Local failure of the rafter

The test results indicate that the supporting walls recorded much higher

acceleration amplitudes compared to the roof structure. Figure 8.56 shows that the

acceleration amplitudes recorded at the top course of the supporting walls were

50% to 63% higher than that recorded at the ridge of the roof structure. This is

likely due to the high stiffness of the roof achieved by introducing the lateral

bracing between the trusses.

8-61

Recorded max. acceleartions at the roof and supporting walls

0.82

1.34

2.81

3.69 3.70 3.70 3.70 3.70

2.62

0.41

0.72

1.18

1.82

1.231.45

2.011.79

1.54

0.00

0.50

1.00

1.50

2.00

2.50

3.00

3.50

4.00

0.05g 0.1g 0.2g 0.3g 0.4g 0.5g 0.6g 0.7g 0.8g

Base motions [g]

max

. ace

lera

tions

[g]

wall at the top roof at the top

Figure 8.56 Recorded max. accelerations at the roof and supporting walls

The response of the roof structure and the supporting walls were almost in phase

except at 4Hz as shown in Figure 8.57 below.

peak acceleartions House 2 at 0.05g

0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

0 1 2 3 4 5 6 7 8 9 10 11 12

Frequency [Hz]

peak

acc

eler

atio

n [g

]

north wall south wall roof Table

Figure 8.57 Natural frequency of the roof and supporting walls at 0.05g

8-64

8.6.3 Natural frequency of the test structure - House 2

The response platform in Figure 8.58(a) and 8.58(b) suggest that in the early stage

of the experiments the wall components were vibrating together almost like rigid

bodies from the top to the bottom courses. The fundamental frequency of the test

structure at base motions of 0.05g was 10Hz, the expected natural frequency for a

single storey conventional masonry structure. The maximum-recorded table

acceleration was 0.65g; the maximum acceleration of out-of-plane walls was 2.0g

recorded at the top of West wall, suggesting that the dynamic amplification at

West wall was 3 times the base motion. The in-plane walls recorded maximum

accelerations of 0.82g and the dynamic amplifications was 1.26 times, about 42%

that of the out-of-plane walls. The natural period T of the test structure at 0.05g

base motions was 0.1sec. It important to note that there was no structural damage

recorded during the 0.05g test series. Tables 8.9 and 8.10 give the response

amplitudes of the test structure at the top of the walls and the roof at ridge level.


0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

1.6

1.8

2.0

0 1 2 3 4 5 6 7 8 9 10 11 12

Frequency [Hz]

peak

acc

eler

atio

n [g

]

Top middle near window middle near corner

bottom roof

a) Out-of-plane East wall

8-65


0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

0 1 2 3 4 5 6 7 8 9 10 11 12

Frequency [Hz]

peak

acc

eler

atio

ns [g

]

middle near corner Top middle near door

b) In-plane north wall

Figure 8.58 Natural frequencies House 2 at 0.05g

The natural frequency of the test structure at 0.2g base motions remained almost

unchanged (about 9.5 to 10Hz) as shown in Figure 8.59. This is likely because

there was no significant structural damage to the test structure during the 0.05g

and 0.1g test series. The out-of-plane wall maximum acceleration was 3.81g

recorded at West wall at maximum table acceleration of 0.95g, and therefore the

dynamic amplification of out-of-plane walls was about 4 times the base motion.

The maximum acceleration of In-plane walls was 2.81g recorded at the North

wall, suggesting that the dynamic amplification was about 3.6 times the base

motion.


0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

4.5

0 1 2 3 4 5 6 7 8 9 10 11 12 13

Frequency [Hz]

peak

acc

elea

ratio

n [g

]

Top middle near door middle near corner roof Table

a) Out-of-plane West wall at 0.2g

8-66


0.0

0.5

1.0

1.5

2.0

2.5

3.0

0 1 2 3 4 5 6 7 8 9 10 11 12 13Frequency[Hz]

peak

acc

eler

atio

n [g

]


b) In-plane North wall

Figure 8.59 Natural frequencies House 2 at 0.2g

At 0.3g base motions, the natural frequency of the test structure decreased to

between 6Hz and 6.5Hz likely due to the progressive damages (Fig.8.60). The

North wall, which recorded early initial structural failure the natural frequency

was much lower (6Hz). The maximum displacement amplitudes for out-of-plane

walls was 10.11 mm recorded at West wall and 3.19 mm for In-plane walls

recorded at South wall. The out-of-plane displacement amplitudes for In-plane

walls were much smaller than that of out-of plane walls likely because of the

direction of the base motions and the roof load. Tables 8.9 and 8.10 give the

response amplitudes of the test structure at the top of the walls and the roof at

ridge level.

House 2 peak accelaeration West wall at 0.3g

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

elea

rtio

n [g

]

Top middle near door middle near corner

roof Table

a) Out-plane West wall at 0.3g

8-67


0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]


b) In-plane North wall

Figure 8.60 Natural frequency House 2 at 0.3g

At 0.4g base motions there was noticeable changes in the response pattern of the

test structure compared to the fist test series. Different positions respond

differently almost in a random manner (Fig.8.61). Depending on the location the

natural frequency of the structure was between 5.0 and 7Hz. The top courses

recorded much higher peak accelerations compared to the courses below the mid

section.


0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]

Top middle near corner middle near door roof


8-68


0.0

0.5

1.0

1.5

2.0

2.5

3.0

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]

middle near corner Top middle near window bottom

b) Out-plane East wall at 0.4g

House 2 peak acceleartion North wall at 0.4g

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

elea

rtio

n [g

]


c) In-plane North wall at 0.4g


At 0.6g base motions test series, the test structure suffered further structural

failures such as partial collapse of the lintel courses and falling out of the window

frame from East wall, widening of the existing cracks at North and West walls to

between 15 to 40 mm, Figure 8.62 suggests that the natural frequency of the test

structure was reduced to about 3Hz as compared to 10Hz in the first series of tests

(0.05g). The response of the walls at different levels was random and the recorded

out-of plane displacement amplitudes at East wall were more than 100 mm, the

reason for such extensive damage.

8-69

At 0.7g and 0.8g most of the monitoring instruments were removed. The structure

experience further destructions by widening of the existing cracks, including

permanent uplift of the base at North wall by 12mm.


0.0

0.51.01.52.0

2.53.03.54.0

4.5

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]

Top middle near door middle near corner roof


House 2 peak accelearation North wall at 0.6g

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

4.5

0 1 2 3 4 5 6 7 8 9 10 11

Frequency [Hz]

peak

acc

eler

atio

n [g

]


b) In-plane North wall at 0.6g


8-70

a) Table accelerations T22

N1 (top) N2 (mid) West wall N7 (top) N6 (mid) East wall


8-71




FIGURE 8.63. MEASURED RESPONSE: HOUSE 2, TEST 22 [0.2g@10Hz]

8-72




8-73




FIGURE 8.64. MEASURED RESPONSE: HOUSE 2, TEST 32 [0.3g@9Hz]

8-74




8-75




FIGURE 8.65. MEASURED RESPONSE: HOUSE 2, TEST 37 [0.3g@ 4Hz]

8-76




8-77



d) Roof accelerations T47

FIGURE 8.66. MEASURED RESPONSE: HOUSE 2, TEST 47 [0.4g@ 3Hz]

8-78

8.6.5 General discussion

From the experiments results in Table 8.11 below one may suggest that the

performance of the plain dry-stack masonry (UDRM) was fairly good compared

to the conventional masonry based on the intensity of the base motion endured by

the system before partial collapse at 0.3g, about 6.2 Richter scale

(Tab.1, Appx.E). If the system is strengthened by reinforcements, is likely to

demonstrate better structural performance under seismic conditions.

At base motions of 0.05g an equivalent of 4.8 Richter Scale or Modified Mercalli

Intensity Scale MMI V (Gendzwill, 1995), House 1 - a plain dry-stack masonry

developed only hair size cracks on the plaster (Tab 8.13) below. An earthquake

of 4.8 Richter scale is categorized as Light earthquake and in conventional

structures, this level of intensity does not cause problems to well designed

structure. Globally it is estimated that more 6000 light earthquakes occurs

annually. See Table 2 in Appendix E.

In the second series of tests using base motions of 0.1g, an equivalent of

5.5 Richter scale or MMI VII scale, House 1, developed diagonal cracks on the

plaster at lintel courses in out-of-plane walls (Tab.8.13). An earthquake of 5.5

Richter scale is categorised as moderate earthquake and it is estimated that

about 800 earthquakes of this level occur every year globally. Moderate

earthquakes cause serious structural damages to weak masonry particularly the

ancient conventional masonry. For example in Morocco, in March 2004 the

earthquake, which left hundreds dead, measured 5.5 Richter scale (Reliefweb,

2004). In Uganda (Okumu, 2005) in 1966 an earthquake of similar magnitude

caused death to 160 people and damaged more than 700 buildings. The

earthquake that completely destroyed a 6-story office building in Welkom (1976),

South Africa measured 5.5 Richter scale.

In the third series of tests where base motions of 0.2g were applied, an equivalent

of about 6.1 Richter scale or MMI VIII scale, House 1 developed some serious

8-79

structural failure at the corners including bending of the door frames (see

Tab.8.13). An earthquake of 6.1 Richter scale is categorised as strong

earthquake, very distractive to conventional masonry buildings and a cause of

death to thousands of people worldwide. It is estimated that about 120 strong

earthquakes occur each year globally. From Table 3 in Appendix E it show that

Tanzania had 4 strong earthquakes and South Africa 2 for the past 50 years

At 0.3g base motion, which is equivalent to 6.2 Richter scale or MMI VIII in

House 1 there was partial collapse of the out-of-plane walls and it was the end of

the tests for House 1. In House 2 there was total shear of the base of the test

structure resulting into isolation of the super structure and foundation.

At 0.6g base motion, which is equivalent to about 6.9 Richter scale or MMI IX in

House 2, there was partial collapse of the East wall; lintel and the window frame

at East wall were ripped off the wall. An earthquake of 6.9 Richter scale is

categorised as strong earthquake, which normally destroys most of masonry

buildings including well-built ordinary structures.

Test results from Table 8.12 and Figure 8.67 below indicate that under similar

base motion intensity the acceleration amplification in dry-stack masonry is lower

than in conventional masonry at higher out-of-plane displacements. This may be

attributed to the flexibility of UDRM. Figure 8.68 below shows that the roof

acceleration amplitudes in House 1 were slightly higher than of House 2 because

of roof flexibility in House 1. The roof rigidity in House 2 is attributed to proper

bracing of the trusses. In both houses test results indicates that the seismic

acceleration increase progressively from the bottom to the top courses.

Table 8.11 Structural damage of the walls at Total failure Out of plane walls In plane walls Base motion Test structure

West East North South

House 1 8% partial collapse

40% partial collapse

Corner failure

(Corner blocks rotation)

Corner failure

(Corner blocks rotation)

0.3g

(6.2 Richter scale)

House 2 major cracks at lintel courses

5% partial collapse

major cracks at lintel courses

Corner major cracks

0.6g

(6.9 Richter scale)

8-80

Table 8.12 Response of the out-of-plane walls and the roof

Responce of the out-of-plane walls

0.0

20.0

40.0

60.0

80.0

100.0

0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35

Base motion [g]

max

. wal

l dis

plac

emen

ts [m

m]

plain dry-stack masonry conventional masonry

Responce of out-of-plane walls

0.0

1.0

2.0

3.0

4.0

5.0

0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35

Base motion [g]

max

. wal

l acc

eler

atio

n [g

]

plain dry-stack masonry conventional masonry

a) maximum displacements b) maximum acceleration

Figure 8.67. Response of out-of-plane walls

Roof responnce

0.00

0.50

1.00

1.50

2.00

2.50

0.00 0.05 0.10 0.15 0.20 0.25 0.30 0.35

base motion [g]

max

. roo

f acc

eler

atio

ns

dry-stack masonry conventional masonry

Figure 8.68. Roof response

Out-of-plane walls

Specimen Base motion

[g]

Max. out-plane displacement

[mm]

maximum acceleration

[g]

Roof maximum.

acceleration [g]

0.05 4.07 1.47 1.01

0.1 9.00 1.52 1.64

0.2 61.80 1.89 1.90

House 1

0.3 90.95 2.09 2.10

0.05 0.97 2.00 0.41 0.1 1.32 2.74 1.14 0.2 2.68 3.81 1.18 0.3 10.11 3.37 1.82

House 2

0.6 132.00 3.88 2.01

8-82

8.6.6 Major observations from the study

The following are the major general observations from the study conducted: -

1. The Shaking Table surface is rigid; therefore it was assumed that the test

would simulate only a very hard soil condition in the prototype.

2. Table motions were applied sequentially, starting from very low to severe

intensity. Therefore the test models were in a weakened condition when

subjected to the more intense base motions.

3. In conventional masonry House 2, the In-plane walls undergo rigid-body

rocking associated with cracking and uplift at the base when the base

motions were above 0.2g (6.1Ricther scale).

4. The large racking (shear) distortions between the roof trusses, observed in

House 1, were significantly reduced in House 2 by introducing lateral

bracing between the two-middle trusses.

5. In House 1 the trusses were orientated perpendicular to Table motion. The

inertia forces from out-of-plane walls and the roof caused lateral bending

of the bottom chord of the trusses.

6. The motions of the UDRM test structure followed the shaking table

motions very closely, particularly at base motions below 0.2g, and the

distortions were generally proportional to, and in-phase with base

accelerations.

7. In UDRM test structure, the out-of-plane walls developed failure by

opening of the perpend dry-stack joints in the centre, two quarters of their

height and developed an out-of-plane hinging action . The failure is

attributed to sliding of the interlocking mid-blocks and rotation of corner

blocks.

8-83

8. In the UDRM test structure, the stiffness of the cracked walls was reduced

significantly compared with the un-cracked state, and the out-of-plane

period of vibration increased significantly.

9. In UDRM test structure, as the progressive damage of the walls increases;

their natural frequency continues to decrease and the degree of response of

the walls varied from one point to another depending on the extent of

damage.

10. House 2, which was reinforced with brick force after every third course,

developed shear at the base and at the lintel level above the openings due

to box action ; or rigid body rotation of the walls.

11. In both test structures the development of diagonal cracks extending from

the window corners was typical.

12. The footing attachment for the model walls was designed to permit both

rocking and up lift; only sliding was constrained.

13. The significant aspect of dry-stack masonry is that the out-of-plane

displacements did not cause failure to the wall by splitting of the units as

in conventional masonry, but rather the failure was by opening of the

vertical joints of the dry-stack interlocking blocks from pull and

bending action .

14. In UDRM test structure, as the base motion intensify, the corner blocks

connecting In-plane walls and Out-of-plane walls undergo rotation

which reduces the rigidity of the structure. The weak corners allow greater

out-of-plane displacements resulting into hinging action at the mid

upper section of the wall and ultimately partial collapse of the walls.

8-84

15. In order to transfer inertia forces developed at the roof level into the

supporting walls, the roof system should be adequately braced in both

orthogonal directions and should be adequately anchored into the

supporting walls.

8.6.7 Recommendations

For further improvement of the UDRM some recommendations are given in this

section. The recommendations are based on the test results focusing mainly on the

physical observations of the structural behaviour of the test model. The

recommendations address the major identified structural problems such as rotation

of the corner blocks and weak corner junctions which were the major causes of

corner failures; excessive out-of-plane displacement of the walls, the major cause

of partial collapse of the walls; bending of the door frames and falling debris.

The recommendations are in two categories. The first category utilise dry-stack

technology solutions, which were initially tested under static loading. In the

second category the solutions are based on the experience in conventional

masonry.

Category I: Dry-stack seismic system solutions

Based on the investigations conducted in Phase I and reported in chapter 7, two

methods are suggested to enhance the capacity of UDRM against seismic loading

as follows: -

i) Introduction of dry-stack lintel beam by constructing the top three or four lintel

courses using conduit dry-stack blocks reinforced with horizontal bars. The

horizontal reinforcements are carefully laid in mortar without interfering with the

interlocking mechanism and anchored to the cast in situ side columns (Fig 8.71

and 8.72c). The static test results of the system reported in chapter seven and

shown in Figure 7.7 indicate that the ultimate lateral resistance of the wall was

15.3 kPa at deflection of 100 mm. From series of tests conducted in Phase I on

8-85

dry-stack wall panels, 100 mm deflection is considered a failure point because as

the out-of-plane deflection of the wall approaches 100 mm due to hinging

action , the interlocking blocks slip out of the interlocking mechanism.

The Shaking Table results for House 1(UDRM) indicate that the peak

accelerations recorded at the top of out-of-plane walls before partial collapse was

2.68g. The magnitude of the Inertia force, which is the product of the mass of the

wall and peak accelerations acting in the direction opposing the acceleration

(Fig.8.69), was about 10 kPa. The partial collapse of out-of-plane walls occurred

at T30 at base motion intensity of 0.3g, suggesting that UDRM system tested will

not collapse if base motion intensity is below 0.3g.

The performance of the proposed seismic system under static load shows that at

lateral pressure of 10 kPa the maximum out-of-plane deflection of the wall was

about 60 mm <100 mm (Fig. 7.7). The performance is encouraging and therefore

the basis for detailed study of the system. It is important to note that the test

structure (Fig.7.6) discussed in chapter 7 was constructed using seismic block

type II (Fig.7.8) and in House I seismic block type I was used, therefore the

comparison made above is very conservative. In order to be able to predict

accurately the amount of reinforcements required for the side colums, it the

opinion of the author that the tests models should initially constructed and tested

first without vertical reinforcements to the side columns. The Shaking Table base

motion intensity should be increased in a sequential manner starting from minor to

very severe while monitoring and recording the mode of failure. The test results

obtained will then form a basis for providing the required amount of vertical

reinforcements.

ii) The second alternative to be investigated is by using the cellular corner blocks,

reinforced and grouted (Fig. 8.72(a)), eliminating the shuttering process. The

construction of the top three or four lintel courses is similar to alternative (i); the

only difference is that the grouted side column cross-section is limited by the

diameter of the block cell, which is 100 mm. Because of small cross section of the

8-86

cell in order to achieve high strength, concrete of high workability should be used

to allow proper compaction. The system should be tested in a similar manner as in

(i) to establish the amount of vertical reinforcements required. In both cases the

possibility of increasing the area of horizontal steel is limited by the size of the

(conduit) cell in the blocks. Experience from Phase I as reported in chapter 7 is

that the diameter of the horizontal bars should not exceed 10 mm to allow

adequate bonding mortar around the horizontal reinforcements. Therefore the

horizontal bars are limited to diameter 6 mm, 8 mm or 10 mm.

The suggested alternative (i) and (ii) should be tested against a test model with

conventional lintel ring beam (Fig. 8.70) for comparisons in order to establish the

advantages and disadvantages of the proposed dry-stack system.

Category II: Adopting experience in conventional masonry

Other proposed options to be investigated in future includes introduction of lintel

ring beam/band and sill ring beam/band as shown in Figure 8.73, introduction of

lintel and sill ring beams casted together with insitu side columns (Fig. 8.74), and

introduction of roof ring beam/band as shown in Figure 8.75. These are some of

options, which have been tested, in conventional masonry for low-rise buildings

in seismic zones in countries such as India Mexico and some of Latin America

countries.

Experience from conventional masonry suggests that the lintel ring beam ties

walls together and create a support for walls loaded along out-of-plane direction

from walls loaded in in-plane direction. The ring beam also reduces the

unsupported height of the walls and improves their stability in the weak direction.

During earthquake shaking, the ring beam will undergo bending and pulling

actions; therefore the main reinforcements should be anchored to the corners

properly. An introduction of conventional cast insitu side columns will enhance

connections between out-of-plane walls and in-plane walls.

In a seismic zone where large wind forces are expected, an introduction of roof

ring beam is suggested for investigations (Fig.8.75). A reinforced roof ring beam

8-87

is expected to provide proper anchorage of the roof structure where the dry-stack

lintel courses with horizontal reinforcements proves to be inadequate. Table 8.14

summarise the above recommendations.

The observations and recommendations given above are based on the test results

of only one UDRM test model and therefore should be considered conservative.

Testing of more than one test models of similar parameters and similar variables

will in future provide more accurate and conclusive results. High costs for setting

up the experiments was the main limiting factor in this investigation. It is the

opinion of the author that the information provided above will be used to provide

a guideline in the development of standard dry-stack system for different seismic

conditions.

8-89

West wall pulls North and South walls while East wall pushes North and South Wall

Figure 8.69 Horizontal Force Distributions

Figure 8.70 Introduction of lintel ring beam/band

8-90

(I) DRY-STACK SOLUTIONS

Figure 8.71 Dry-stack lintel beam

(c)

Figure 8.72 Insitu side column

8-91

(II) INTRODUCTION OF CONVENTIONAL METHODS IN DRY STACKING

Figure 8.73 Introduction of lintel and sill band

Figure 8.74 Introduction of the bands and cast insitu side columns

Figure 8.75 Introduction of Roof band

th chapt 8 final

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