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FINAL REroRr '!'HE APPLICABII1ITY OF '!'HE WFAP-86 PIIB CAPACI.'lY

imDlcrICN MEIH:>D :roR OONNECI'IaJr SOlIS ani

BEHAVIOR OF A SHEEI'PIIE WAIL FOR STABILIZATICN OF A SLOPE IN :EORl'I.AND

Project 86-7 by

Richard P. ~, Professor Kenneth R. Demars, Assoc. Professor William A. Shaheen, Grad. student Viroj Mekaga.roon, Grad. Student

JHR 88-181 octdber 1988

'!his research was sp:>nsored by the Joint Highway Research Advisory council of the University of cormecticut am the COrmecticut of the University of cormecticut ani the Connecticut Department of Transportation, am was carried out in the Civil Engineerirg Department of the University of Connecticut.

Table of Contents

I. WEAP.86

Introduction •

Scope of the Research

Background •

The WEAP-86 Program

Failure criterion

Analysis •

Conclusion •

II. Portland Embankment Stabilization

Introduction •

Background •

Instrumentation Program

Result.

Conclusion •

1

1

4

12

15

20

33

36

36

38

41

52

1

Introduction

The engineer requires reliable methcx1s for predicting the

capacity of piles as driven in the field. Early approaches used

semi-empirical methcx1s for making these predictions. The success of

these early attempts was mixed. The Federal Highway Administration

sponsored research to develop better methcx1s for predicting pile

capacity and a method evolved based on the work of Smith (1962) using

the wave equation analogy as a means of modelling the pile driving

operation and the resulting capacity of the pile. All mathematical

models must be calibrated to field conditions in an area. The

Connecticut Department of Transportation has conducted many load

tests on piles over the years. In this phase of the project, pile

capacity as predicted by WFAP was compared to the capacity measured

by field load tests.

Scope of the Research

This phase of the research conpares the pile capacity predicted

by the WFAP-86 version of the Wave Equation Analysis of piles (FHWA)

to available appropriate pile load test results. Included in this

project are the comparison of capacities for both monotube piles

which develop their resistance through skin friction and H-Piles in

glacial till whose capacity appears to be end bearing. The Soils and

Foundation Division of ConnIX:1.[l made the results of 240 pile load

tests available for comparison. These tests were screened for

suitability for this investigation and many had to be excluded. The

most common reasons for excluding individual tests from consideration

were: The pile was not loaded to failure or that the pile was driven

to bear on rock.

2

Field load tests are often "proof" tests to demonstrate that the

pile as driven, can withstand at least twice the design load without

excessive settlement. In these tests the ultimate capacity is not

determined and a comparison with a predicted value from WEAP can not

be made. When a pile is driven to rock, the load test shows

primarily the elastic shortening of the pile and yields no useful

infonnation on the soil pile interaction properties.

After reviewing available test data, seven steel H-piles. ranging

in size from liP 10 x 42 through liP 12 x 74 were selected for

comparison. 'lhe test data for the seven selected monotube piles

indicated that their ultimate capacity depended on skin-friction

together with an end-bearing component. Comparisons of field tests

and WEAP were made for these piles.

An example of the load tests used in this investigation is shown

in Fig. 1. 'lhe pile shown in Fig. 1 is an end-bearing pile driven

into glacial till. The pile size is liP 12x7 4.

txj ~.

1 .... .

.r--.

rOWN J-!,ell?iI='"OR.J> PROJECT .:r- 84-PROJECr NO. 6 '3-/37 SOILS ENG'R. ~p.M.

srRucrURE hlB. ~~ __ ... $USSrR. UNJr _.p.~t::. =- - .jg' ...... __

DArE DRIVEN 16 4pr "a rEsr PERIOD _~ I"1AR- ~J9M ~2.

Pi..'" #z _'

r­,..,........."

PILE rYPE Go H. Jlt'l!:.~.t. ... ~I-1'· 'PILE SIZE 11 WEIGHr .L2.8P74 DRIVEN LENGrH~...!...:.~!.' EXlsr •. GR. ELEV. 3 ?o.' rIP ELEV. _.1:.0'._ ....

HAMMER MODEL _~L_Q.6: ___ _ RAM WT. a STROKE .G,)..QQ~ - 3:.." II

ENERGY (ACTUAL) _j~SO.Q#::/~ 8LOWS/ MIN. (ACTUAL) ... ,.,_,, __ .. __ .

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REMARKS: __ ';;:0/:; ~ara. ~"., ~~ d: B4s a~.t)i' (t(_~_.&~( /?/Ic:: . __ ~tt.cdz()a - J ct.~-A14",.sf lP~.t.~~ . --,--t:=- - - . 7V1'S!''1,,-L\SI.':\'''''':'

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4

Background

Pile f01.U1dation design requires prediction of both pile capacity

and the length of pile. Prediction of the ultilnate bearing capacity

of pile may use two approaches; static analysis and dynamic

analysis. '!he static analysis estilnates capacity from soil

properties with little regard to the manner by which the pile is

placed, 'Whereas the dynamic approach uses cons&Vation of energy

principles without regard to soil properties and behavior. Each

approach has inherent advantages and disadvantages.

'!he static analysis requires accurate soil strength, soil unit

weight, pore pressure information, and lateral earth pressure

coefficients, together with pile dimensions to accurately predict

pile capacity. '!he soil information is usually estilnated. fran the

boring logs. '!he problems involved with static analyses have led to

the widespread use of simplistic energy-based fonnulas 'Which do not

consider soil properties in their development.

Dynamic analyses are based on equating the energy transferred to

the pile by the hammer to the work done by the pile in penetrating

the soil a finite distance. '!he penetration distance is called the

set. '!his approach is attractive since the set can be obBel:Ved in

the field and used to control capacity. However, the physical

properties of the pile and properties of the soil are not considered

in early dynamic fonnulations. '!he first such fonnula appeared in

Engineering News Reco:rd magazine and hence is known as the ENR

fonnula. '!he original ENR equation to calculate the ultilnate pile

load (Pu) as a function of the final pile set is given by:

Pu = Wr h s+l

where: Wr = weight of ram (kips)

h = height of fall (inches)

s = final pile set (inches)

5

'!he nominal factor of safety for this approach is 6 but field tests

have frequently shown that the actual factor of safety by this

equation can be less than 1, or greater than 6. Even though the ENR

fonnula is one of the most unreliable methods available to predict

Pu, it is still widely used. to control pile driving due to its

simplicity.

Soon after the development of the ENR fonnula, many others

developed dynamic fonnulas for predicting Pu. Among the available

fonnulas, the Hiley equation, and the Rational pile fonnula appear to

be most accurate. '!hese equations were the first to introduce a tenn

to account for soil stiffness. '!he Hiley equation is shown to be,

Pu = eWrh s + (IG. + K2 + 10)

Where: e = ~ efficiency

Wr = weight of ram

h = ram fall

s = final pile set

Wp = weight of pile

n = coefficient of restitution

kl = capblock and pile cap stiffness

k2 = pile stiffness

k3 = elastic compression of soil

6

'!he k3 tam is a first step fran the early dynamic formulations

to a more COlTp1ete model. '!he Hiley and Rational formulas were the

first to consider soil stiffness in an energy formulation.

Application of the wave equation to pile driving yielded a model that

allows for pile dimensions, variable soil stiffness along the pile

profile, soil and soil damping, system losses, etc. This was to

become known as the wave equation analysis of piles (WEAP).

rrbe wave equation was first, put into practical fonn by E.A.L.

Smith1 in 1962. rrbe greatest advantage of the wave equation is

that it attempts to account for variation in a large number of

physical parameters for any particular pile driving case. It

predicts ultilnate pile capacity as a function of the final blow

count, and estilnates stresses in the pile that develop during

driving.

The wave equation uses physical properties of the pile soil

system, accounts for energy losses in the system; and in-situ pile

driving infonnation to predict static pile capacity.

rrbe wave equation may be derived fran consideration of the

intemal forces and motion produced on a segment of a free1y­

suspended elastic bar subjected to an inpact. at one end. rrbe

resulting 1-0 equation is:

where c = longitudinal wave velocity

U = longitudinal displacement

x = longitudinal coordinate

t = time

This one-dimensional wave equation has been derived fran the

equations of motion as outlined in Ko1~, and has been solved in

7

closed analytical form by ~ for a homogenous 1-0 bar

embedded in an homogeneous elastic medium. standard linear,

isotropic, elastic material properties are assumed for Pochh.aJmnerI s

exact solution. In practice; however, viscous soil damping, variable

pile cross-sections, and variable pile cushion stiffness

considerations all make exact analytical solutions impractical if not

impossible to solve in closed form.

In addition, for a pile with non-zero skin friction, the

resistance of the surrounding soil must also be considered.

Modifying the above equation yields:

d2U ? d2U dt2 = c- dx2 +/- R , where R is a soil resistance term.

'!he WFAP-86 computer program uses a finite-difference numerical

solution to solve the above modified equation.

'!he approach used by WFAP-86 was first developed by Smith to

determine the pile-set (or b101f1 count) for a given ultimate static

pile load. '!he pile system is idealized as shown in Figure 2 and

consists of the fo1101f1ing elements.

1. A weight (ram) to which an initial velocity is imparted from

the hannner.

2. A capb10ck (hannner cushion)

3. A pile cap (anvil)

4. A cushion block (for concrete piles)

5. '!he pile

6. '!he supporting soil

(A) ACTUAL .SYST.EM. DIESEL

AIR I STEAM

.. "'~I---- RA M --a ..

o - ANVIL---~~~ ~----r' CJ;n ' CAPBLOCK

::==:~ . LJ '1.I~HElMET --.. -.... -:..----.. ~., :t:l~ C·USHION~-~~ .

·o~ PILE -----..

(C) SOIL RESI'STANCE: ' o :::E ~ z >­o a: "'--___ __

VELOCITY

8

-':.,.:;':"', ,

:'f

;,.. ', - .::~~ .... ,;, ... : . . :"; ., . -• .' _ ~ ;f. ~~. . .... ~

. ~ .i,·r:-: . .. : .... ':-:' / . . :t ~~ · f: " I~ > : .

-...... '. .' .. -, .:., ~ " \, - . ~ . .. ',"

I I .' .

~' . ...

DISPLACEMENT

Figure 2. Actual Soil-Pile System and Idealized Schematic Model

'!he ram, capblock, pile cap, cushion block, and pile are

represented by appropriate discrete weights and springs. '!he

frictional resistance on the side of the pile is represented by a

series of springs and dashpots, while the point resistance is

modelled by a single spring and dashpot.

9

Soil does not follow the constitutive laws for an isotropic

linear-elastic material by being highly nonlinear for laI:ge strains

and. able to defonn plastically as well. Smith I s model of the

constitutive laws of a typical soil subjected to a stress reversal is

shown in Fig. 3.

Load

Ru(m)JY(m.t )

·.~u(mj· -

D<zformation

Ru(ml·

D

Figure 3. Typical Response for Soil Undergoing a Stress Reversal

10

The loading path OABCDEFG represents loading and unloading in

side friction. For the pile point, only compressive loading is

considered and the loading and unloading path in OABCF. The Q(m)

tenn, called the quake, is defined as the maxllm:nn "linear"

defomation the soil may undergo before the onset of plastic

defonnation. A idealized load-defonnation diagram such as in Figure

4a may now be established separately for each spring, so that

K' (m) = Ru(m)

Q(m)

where K' (m) is the spring constant during elastic defonnation for the

extemal spring (m). To aCCOlmt for the effects of dynamic loading

during pile driving in increasing the instantaneous resistance of the

soil, a Kelvin-type rheological model is used (Figure 4b). In the

Kelvin model, the damping constant, J, represents the an additional

resisting force which is proportional to the velocity of loading.

Load

E

Friction link limits spring load.

Spring constant K'

Darormation

RJm)

o

Soil Rllsistancll \ R

..... _ ........ ___ . ---.J .=TI.isPlacamant D

Damping constant J.

11

Figure 4. Idealized Elasto-Plastic Stress-strain Behavior of

Soil (a), and Visco-Elastic MOdel (b)

'!here is considerable uncertainty regarding the selection of

quake and damping values to be used in a WEAP analysis. However; a

12

commercially available electronic pile driving analyzer, using a

sampling probe attached to the pile butt during driving, can monitor

and detennine in-situ quake and damping values. 'lhese values

together with a computer program, called CAPWAP4 , further refines

WFAP prediction accuracy.

'!he WEAP-86 Program

In order to model and predict pile load capacity using WEAP-86,

the soil and pile physical properties must be input to the computer

program via a data input program called wa6-IN. Required physical

properties include pile driving hammer infonnation, pile and helmet

cushion weights, lengths, and stiffnesses, pile-segment

discretization properties including masses, damping, and stiffnesses;

and soil-pile damping and soil quake values. The WFAP program

provides default values for the soil input parameters for toe and

side quake tams (.1 in.) and toe and side damping tams

(.1 sec./ft.).

The material properties required by the WEAP-86

finite-difference ccxie include the modulus of elasticity of the pile

helmet, pile segments, and pile cushion. In addition, assmned

skin-friction distriliution schemes and end-bearing/skin friction

percentages, soil quake values, and damping values all must be

specified. A sample input card is shown in Figure 6 which describes

the soil-pile system of Figure 1.

type II.Out 1

q~'~\j ECHO PRINT OF INPUT DATA

o " 1.200 .000 .0 .000 .0 .000

30.500 21.160 30000.l')OO VULCAN VOL 06 3 1 0

6.5000 39.19(~ 17.4q(~

. 00(1 • 000 • 8(10 2.~~0 2.350 .000 • (le)(I(I • (11)00 • (II)(H)

• 1 00 • 1 00 • 1 ()(I

0 ··0 0 .000 .500

492.000

3.0000 .010

21709.8 • ')0('(1 .100

ll~.O 150.0 2~0.0

.0 .0 .(1

15 0 .800 .010 .8'W

3.0000 2

21709.8 .0000

6 0 0 .010

.0 .010

.6700

.0

.0

WEAP8o: WAVE EQUATION ANALYSIS OF PILE FOUNDATlONS 1986, VERSION 1.(101

hpilf.! pg 1::'1 I

HAMMER MODEL OF; VUL 06 MADE BY: VULCAN

0 0 .0

.0

ELEMENT WEIGHT STIFFNESS COEFF. Of" D-NL. CAP DAMF'G

)

(

(KIPS) (KIlN) R£ST! TUnON FT H(lFT/S) 1 6.500

CAP/RAM 1.200 177183.5 .8l~ .0100 26.3

ASSEMBLV WEIGHT STIFFNESS COEFF. OF D-NL. ! (KIPS) (KIlN) REST !TUnON FT

1 2.350 21709.8 2 2.350 21709.8 .800 .0100

HAMMER OPT I DNS I HAMMER NO. FUEL SETTG. STROKE OPT. HAMMER TYPE DAMPNG-HAMR

206 1 0 3 2

HAMMER PERFORMANCE DATA RAM WEIGHT RAM LENGTH

(KIPS) (IN) b.50 ,!.9.19

RTD PRESS. (PSI)

.00

ACT PRESS. (PSI)

.00

MAX STROI:E (FT> 3.1)"

EFF. AREA ( IN2)

.00

STROKE EFFICIENCY 1FT)

3.00 .67(1

IMPACT VEL. (FT/S) 11.37

WEAP OF 1986 hpile P9 75 I

PILE PROFILEr LST \ AREA E-MOD SP.W. WAVE SP EAIC

(FT) ( IN2) (KS1) (LB/FT3) (FT/S) (K/FT/S)

.00 21.8 30000. 492.000 16806.8 ;sa. 8 ::SO. 50 21.8 30000. 492.000 16806.8 38.8

WAVE TRAVEL TIME - ·2L/C - • 3.629 MS

PILE AND SOIL MODEL FOR RULT .. 10CI.O KIPS

0

NO WEIGHT STlFFN D-NL SPLICE COR SOIL-S SOIL-D QUAKE L 8T

2 4 6

TOE

(t<IPS) (KIlN) (FT) .378 lu702. .010 • 37e 10702 • .(ltO .378 10702. .010 • 378 10702 • .010

PILE OPTIONSJ N/UNIFORM AUTO S.G.

o o

SOIL OPlIUNS: X SKIN FR % END BG

15 85

(FT> (KIPS) (SIFT} • (lOCI .8S0 2.5

-1. (IOu 1.000 2.5 -1.000 1.000 2.5 -1.000 1.000 2.5

8:1.0

SPLIC£S DAMPNG-P

DIS. NU. 5 DAMPING 6 SMln~-1

• 100 .100 .100 .100 .100

D-P VALUE O"-;/FT IS)

• 777

ANALYSIS/OUTPUT OPTIONS:

<IN} (FT)

.100 5.08

.100 10.17

.100 20.33

.100 30.50

.100 f l

ITERATNS DTeR/OT<X) RES STReSS lOUT (I

AUTO SGMNT OUTPT INCR o 160 (\

.0

AREA (INU2) 21.8 21.8 21.8 21.8

MAX T (MSl (I

13

Figure 5. WFAP-86 Input card for Pile Installation in Figure 1.

14

Of considerable interest to the geotechnical engineer are the

quake and damping values. These are very difficult to estimate from

available soil properties, but experience over the years suggest

certain guidelines. Forehand and Reese5 have compiled some

errpirical quake and damping values ' as shown in Table 1. Unless

otherwise stated, WFAP-86 provides a default value of .1 in. for toe

and side quakes (Smith), and .1 sec./ft. for toe and side damping.

'Ihese values are the recanunended default values for using WFAP-86,

unless values from any other rational basis, such as the

CMWAP/dynamic pile analyzer, are available.

Table 1. Empirical Values of Qr, or, and Percent Side Friction5

Soil .I(p) ·. (sec/ft)

Coarse sand 0.10 0.15 Sand gravel mix~d 0.10 0~15 Fine sand 0.15 0.15 Sand and clay . or .loanl', ::at leastSO% ,·/ 0.20 ;' :; 0~20," .

Q(pn. e.in .. ·san.d " . . Sitt and fine sand ,

: underWD ·~t·hai'd strata

,~ . , · ·~tt.~4~-·~.ii~'~.1s~.~· . strata

Side _ . Adhesion (%"ofRU> ..

35 75-100

100

15

Failure Criterion

lJhe tenns "ultimate load" and "failure" for a pile re<;;lUire some

discussion. As stated by Peck6 et. al., the ultimate or failure

load for a friction pile may be constructed graphically s:iJTq;>ly by

extending the two lines obtained from the load vs. settlement cm:ve.

For a friction pile these two lines are well-defined. HO\'Never, for

piles that develop significant end-bearing resistance, the load -

settlement cm:ve increases monotonically as the load increases due to

the skin friction being mobilized first, then the point resistance.

Van der veen7 has postulated that the load settlement cw::ve is of

the general form P = Pu(1-ebx), '\Nhere x is vertical settlement and

b is a constant determined by a least square cw::ve fitting scheme.

Van der Veen's criterion defines the failure load as the load

value corresponding to the vertical asymtote of the load vs. butt

defomation cw::ve. Van der Veen's criteria are shown schenatically

in Fig. 6. A least square cw::ve fitting computer routine was

developed to find Van der Veen's (VDVM) ultimate load from available

load - settlement infomation. However, VDVM indicates excessively

large pile capacities for some data. 'Iheoretically, Pu will occur at

an unbounded displacement value using the VDVM criterion.

16

Load Pu (Van der Veen)

~--====~------------------------------------- y

x

. ,.

Assume of form

y

ot

A x

=

~

> >

-<xx A( 1 - e )

@

@ @

Figure 6. Van der Veen's Approach for Detennining Pu.

Alternatively, one could use Terzaghi's criterion8 which

defines P ult. as the load at which pile settlement equals 10% of the

pile base diameter. Hawever, this approach fails to consider the

load - settlement curve of the pile and the associated. soil-pile

behavior and may also lead to somewhat high Pu values.

Another unpublished failure criterion, developed by Q'Nei119

plots the time rate change of settlement against applied load. The

17

generated curve consists of two lines, 'Whose point of intersection

corresponds to the failure load as described by O'Neill (Figure 7).

In o:r.U.er to use a 'Neill's criterion, time vs. settlement curves for

each load increment must be available. since this criterion has only

been presented recently, the CDNNDJr data base does not contain the

necessary info:t".ltlation.

Wi dt

Time tl t2

~~------------------------------------------~ t

s

Slope = ds PI dt

P2

P3 I increasing slope

Pu (O'Neill) o~ ____________________ ~~~~ ________ ~

o LOAD

Figure 7. O'Neill's Approach for Detennining Pu.

18

A useful definition for the failure load of a pile could be the

point of maximum curvature on the load-settlement curve. 'Ibis point

represents a finite settlement beyond which the settlement per load

increment increases. The point of maximum curvature can be

detennined by eyes.

An example is shown in Figure 8. The degree of curvature is

mathematically equal to the 2nd derivative of y with respect to x,

and in words represents the rate of change of slope with respect to

x. However, for practical purposes, one can locate the point of

maximum curvature (minimum radius of curvature) by inspection.

Load Pu (m.c.)

o~~----------------------------------------------. y

x

~---------- Min. radius of curvature located here by inspection.

Figure 8. Maximum CUIvature Approach for Determining Pu.

19

'!he concept of the maximum curvature criterion is along the same

lines Peck's criterion for friction piles.

Figure 9 shows a typical load vs. butt settlement curve and a

comparison of ultimate capacities obtained from using the VDVM,

Terzaghi, and maximum curvature failure criteria.

o

Pu Pu Pu Load (Me.) (VDVM.) (Terzaghi) ~O~ ______________________________________________ ~~ y

x

Pu (Terzaghi) is D at 19% 9.l*B'= D @ Pu(Ter)

H-Pile 12*53 EBT. (page.75)

Figure 9. Pu as Obtained by Three Methods

20

PILE ANALYSIS

The follow'ing OONNOOI' end-bearing H-pile and friction monotube

pile installations were modelled using WEAP-86 and the program

default values for quake and damping. Their actual measured capacity

from OONNIXYr load tests are shown in tabular fonn in Table 2 as

inteJ:preted by tvIo methods: VDVM and maximum cu.:rvature.

Table 2. :Physical Characteristics of piles studied.

Table 2

Pile capacity Prediction :ey

IDeation Pile VDVM. Max. AASHIO. WEAP.86 Note Size CUrv.

I-84 steel H 352 426 151 210 15% Skin #7 Hartford 12BP53 Dist., Smith Pier#4-WB wr = 0.75

I-84 steel H 282 250 135 450 25% skin #3 Hartford, l2BP53 Dist., Smith West - wr = 0.75 Abutment

I-84 West Steel H 318 270 140 305 15% skin #6 Hartford, 12BP53 Dist., Smith Pier 1 ST = 0.75

I-84 Steel H 278 190 124 180 10% skin #2 Hartford, 10BP42 Dist., Smith

wr = 0.75

I-84 Hf Steel H 318 220 131 130 25% skin #6 OVer 12BP74 Dist., Smith NY.NH. & wr = 0.75 HRR.

I-84 Steel H 356 300 175 405 5% skin #5 Hartford 12BP42 Dist., Smith E.B. Road wr = 0.75 Way

I-84 Steel H 280 222 117 140 5% skin #7 Hartford, 10BP42 E = 1450 psi.

den = 75 pcf

21 Table 2. (Continued)

Pile capacity Prediction By

location Pile VDVM. Max. AASHIO. WEAP.86 Note Size CUtv. DFSIGNIDAD

1-84 Precast 502 400 227 275 10% skin Hartford, Con. 24" #7 Dist., S-E Rdwy. Sq., Type wr = 1.0 Viaduct III tip.

1-84 Timber, 42 18 13 9 60% skin Nook - 12" dia. #7 Dist., FannSewer III tip. E = 1450 psi.

Derby, Concrete 222 156 75% skin Rt. 34 Monotube #2 Dist. Bridge over wt = 0.75 Naugatuk R.

Graton Concrete 250 180 125 151 50% Skin, Southington Manatube #3 Dist. 1-95 wt = 0.75

Coventry Concrete 304 264 144 168 75% Skin, Monotube #2 Dist.

wt = 0.75

Farmington Concrete 162 124 50 133 60% Skin, Manotube #7 Dist.

Coventry Concrete 212 150 105 178 75% Skin, 1-84 Manatube #2 Dist.

Derby, Concrete 224 156 85 75% Skin, Rt. 34 Manotube #2 Dist. Bridge over Naugatuk R.

-------- -------------------------------------------------------------------------------------------------------

22

It should be reiterated that the two program input parameters

with the greated uncertainty are the soil quake terms and the soil

damping terms. other input terms such as the mod.ulus of elasticity

of steel, unit weight of steel, length of pile, and to a somewhat

lesser extent, the skin friction distribution, hammer energy

transferred to the pile, and cushion stiffness are known to a

relatively high degree of accuracy.

'!he major focus of the study deals with the effect of the

selected quake and damping terms upon the predicted pile capacity for

end-bearing H-pile capacity; in addition, a mnnber of computer

modelled monotube piles considered the effect of different skin

friction distribution patterns as well as selected quake and damping

valve effects. WEAP-86 predictions were generally conducted by

varying only a single parameter at a time so that only the effect of

that particular parameter is noted.

a) End Bearing H-piles

A total of 7 steel H-piles driven into sand/glacial till were

selected for study based on availability of appropriate load test

infonnation for pile tested to failure. '!he required infonnation

includes soil boring infonnation and blow count data taken during

driving. Data from the soil borings are used to estilnate the

fraction of the ultilnate load carried in side friction as well as the

distribution of stresses along the pile shaft. WEAP program runs

were made using the physical properties of the hammer used, the type

of pile driven and the bearing soil. '!his phase of the analysis used

the program default values for the quake and damping (smith type)

terms. '!he results, shown in Table 2 and plotted in Figures 9a and

9b, suggest that the standard default values of quake (Qr=.1 in.)

23

and damping (DI'=.1 sec./ft) for end-bearing H-piles statistically

undet'predict measured pile capacity as intel:preted by either the Van

der Veen or maximum curvature failure criterion (Figs. lOa and lOb).

GOO

500

400 " . i' S:: '" J 300

3 200

100

o

WEAP.86 (Default OT = 0.1 DT - 0.1) Max. Curve.. Foilure Crfterfon

PeR. != 2 1/ V --

Va 1/ -_. / :/ ...

14° I~J ...

tJ

I I;: ..

kV i

o 200 400 eoo lOGd Predtcted by WEAPM (J<tpa.)

- FS. - 1.0

Figure 9a

4 I i

800

,.... 1-....,

! !

500

400

300

200

100

o

WEAP.86 (Default OT = 0.1 DT - 0.1) \'DVM. Fctlure Crlterton

// ~

..

+1 + /t J++ / +

...

/ 7 1/ i

o 200

Load Pred1cted by WEAP_80 (KIps.) - F& .. 1.0

Figure 9b

Pile Capacity Ratio & Freq.of Occurence

24

----

VDVM. Criterion 8~--------------------------.---------________ ~

0.2 - 0.4 0.6 1.0 1.2 1.4 1.6

PIle Ccrpo1\y Ratto .. (pu)*eap/(Pu)vdvm.

Figure lOa

t I

" j C

t i ~

J I ]

Pile Capacity Ratio & Freq.of Occurence Max. CUMt. Cnterlon

8~-----------------------------------------------

1

180

170

180

150

140

130

120

110

1C)()

90

80

70

eo 80

0.2 - 0.4 OA o.s 1.0 1.2 1.4 1.6 1.8./

Pile Copoo~ Ratto .. (Pu)weap/(Pu)mc.

Figure lOb

Monotube Pile: C.I.P. Concrete Load PredIcted By WEAP86 W/RSA &t WOjRSA

--+--"-+--/-1~

+/ I //

,.-+

80 100 120 140

Load Predfct.d By W!APM WO/RSA' (kfpe) - FS. .1.0

Figure 11

1to 180

26

b. Monotube piles

Seven monotube pipe piles were also modelled using WFAP-86.

Monotube piles are nonnally used as friction piles. '!he effect of

the assumed skin friction distribution along the pile shaft requires

consideration as to the residual stresses, damping, and quake

values. The i.Irportance of residual stresses developed in driven

piles has been demonstrated by HollowarQ• The compression of the

pile and soil on the downward stroke when the ram rebounds results in

the retention of compressive forces in the lower part of the pile,

and these appear to depend on the soil-pile interaction only,

independent of the impact driving apparatus used. The magnitude of

the residual loads increase as the axial stiffness of the pile

decreases. Impact driving of a pile with low axial stiffness will

cause the pile to compress in the vertical direction and expand in

the radial direction. As the hanuner rebounds the pile attempts to

return to its orginal length, but the soil along the pile resists

this expansion through friction. This soil-pile interaction in the

pile results in the development of a residual compressive force at

the pile tip which remains even without an applied force at the butt

end. When a residual load remains after driving, a portion of the

pile's end-bearing capacity has already been mobilized as noted by

Braud et ale ColWentional ·pile instrumenation are zeroed at this

time, ignoring any residual stresses in load test interpretation.

The actual end-bearing capacity of the pile is the measured value

plus the residual value. While the effects of residual loads are not

measured, failure to account for residual stresses in the WFAP

analysis leads to low predicted pile capacities (Fig. 11).

27

Monotube piles - Analysis

Seven monotube pipe piles bearing in sand were also modelled

using WEAP-86. The results, shown in Figures lOa and lOb, suggest

that the standard default values of quake (QI'=.01 sec.) and ~ing

(o:i:=.l sec./ft) together with use of the R.S.A. for monotube piles

statistically underpredict measured pile capacity as interpreted by

either the Van der Veen or maximum curvature failure criterion.

Sensitivity Analysis

A parametric sensitivity analysis was conducted to investigate

the effects of pile-soil parameter variance on H-pile and monotube

predicted capacities.

H-piles- A series of WFAP pile capacity prediction runs were

made capacity predicitions were made by varying the toe quake from .1

in. to .05 in. and by varying toe ~ing from .1 sec./ft. to .05

sec./ft. The variance in a given parameter was conducted while

holding the other constant in order to note the effect of varying the

parameter of interest only, holding all others at default values for

Figs. 13 and 14. Fig. 15 shows the effect of varying Qr and Dr at

the same time. Decreasing either parameter by 50% resulted in an

increase in predicted pile capacity of about 15%. Figures 16a. and

1Gb. show that for QI'=O.08 in., for example, the composite actual

capacity versus predicted capacity value reflect this increase in

predicted capacity by a shift of all points to the right. other Qr

and Dr values selected (See Figs. 17a, b; 18a, b) show similar

trends. A major result from this parameter variation study is that

the net accuracy of the solution remains essentially unchanged

regard.less of Qr or or, provided values near WEAP-86 recommendation

are selected. Hence, predicted pile load capacity is not highly

Monotube PiJe : C.I.P .. Concrete 28 W'i.AP.88 WIth and Without RSA.

180 -

110

" It

~ 160

"" ~ 150

~ 140 ~-

Q) GO r1. 1~

~ 120

~

J 110

t 100 0.

'8 90 .c3

80

/ :/ +

/' + "'

/~ +

// :/

:/ +

:/ :/

V. +' I

70 80 100 120 140" 180 180

Load Predtoted tJy WE'AP.8fS w/RSA.(Jdpe..)

Figure 12

WEAP.86 At DT - 0.1 Pile HP 12-53 EBl'.

250

240

230

220

~ 210

~- 200

~ 110

I 180

'R 170 0.

] 180

150

140

130

120 0.0& 0..1$ 0.25 0.35

QT. (tn.)

Figure 13

WEAP.86 At QT = 0.1 PIle HP 12'U l1li'. 29

~~~------------------------------------~

1. CI.04 0.1.

Figure 14

WEAP .86 At DT - QT -PIle HP 12*&3 EBT.

210

210

2-tO

; UO

• 220

I 210

1 200

110

1.

110 0.04 0.1.

Dr - QT (In.)

Figure 15

-" 8-l "V

~

400

~

200

+

0.40 0.80 0.80 (ThouecftcSs)

Lood Predtcted by WEAP.88 (Ktpe.) ---..- FS. .. 1.0

Figure 16a

1.00

WEAP.86 (Default OT = 0.08 DT = 0.1)

30

1 .. 20

ASSHTO. Max. Curvature cnt.rton

eoo T~- r I r T--r I -,-

t- ~ --t-~' ~ l~_I~-I---"~r' ~-+-i~-f ~ :~~~~!~~ ~- +J

200

0.00 - 0.40 0.80 0.80

Load Predtcted~~~ (Ktpa.) - FS. .. 1.0

1 .. 20

Figuie 16b

"

'" ~ j ]

600

100

o

WEAP.86 (Default QT = 0.05 DT = 0.1) VDVM. FGtlure Crfterton

I I IV I !

_t---_-t-· t-~-- ____ I ~.

_._,,~~~. I

t·~-r.-· ---

/ o

1/1

+ I +

+ .

+

~ -I

"":::ror..,.,,...~. r---

/ i

200 400 eoo 1.oOd Predfcted by WFAP..80 (KIPI-)

- FS." 1.0

Figure 17a

.

WEAP_86 (Default OT = 0.05 DT = 0.1) ASSHTO. Max. Curvoture Crfterton

I

31

-~

f500 -----r------r--l -T-- r "--""'--~I -----.----.-.--r\----,

I! i! I !

400

300

200

, I I

++ '-- ---t :--r-t-­-+----+----+---1 ---Y--::! +~t-+- I t·-

o~--~--~---+--~--~----~--~--~--~ o 200 400 eoo 800

load Predtcted by WEAP.8e (Kfpe..) - Fl. -1.0

Figure 17'0

400 '" .-2-.....,

J 300

"0

~ 200

400

t-g 11 3QO ~

] 200

100

Q

o

WEAP.86 (Default ·QT = 0.1 DT = 0.05) vovu .. FaUure Crtterion

+

200 400 600

Load Predtcted by WEAP.8& (Ktpe.) - FS. .1.0

Figure l8a

I \

800

WEAP.86 (Default OT = 0.1 DT = 0.05) ASSHTO. Max.. Curvature Criterion

I / ! ~ I

i ,: j

i

I IL I -~.-+-

t f---- ----1------- --f----

I I I I

ti/ . • -T

+ +

/ I

/ i

i •

32

-_.-

o 200 400 600 800.

Load PredIcted by WEAP.18 (Ktpe.) - FS.. -1.0 Figure l8b

33

sensitive to large, +1- 100%, changes in the quake and damping values

for end-bearing H-piles.

Monotube piles A series of WEAP pile capacity prediction runs

were made by va:tying the toe quake from .15 in. to .05 in. and by

va:tying toe damping from .15 sec./ft. to .05 sec./ft. Results

(Figs 18a, 18b) show that a total variation of 15% occurs over the

entire range of selected quake and damping values. Next, both

parameters were dried to note the composite effect of va:tying or and

Pr at the same time (Fig. 19). Finaly, variation of skin-friction

percentage as a fraction of total variation of about 15% occurs from

0% to 100% skin friction assumed of the pile shaft (Fig. 20).

COnclusions

The wave eq:uation analysis of piles (WEAP-86) finite difference

computer solution package applied to end-bearing H-piles and

friction-type monotube piles in sand show the following trends.

1. Selection of quake and. damping values do not appear to greatly

influence predicted pile capacity (± 15%).

2. Percent skin friction assumed for monotube piles changes

predocted capacity about 15%.

3 • Residual Stress Analysis should be used when modelling

monotube-type piles.

4. Default values of quake and damping yield predicted capacities

that are statistically lower than the measured capacities.

5. WEAP-86 provides the user with some insight into the pile

driving process and. effects of various properties on the pile

stresses and. capacities.

Monotube Pile: C.LP. Concrete 34

WIth RSA.

~r-- l 95 "-,

94 ~, 93 .'

92

91

90

89

"'" 88 .; ~ 87 'V ae :t 0- eo

84 83 82

81

80

79 78 17

0.05 0.07 0.0& 0.11 0.13 0.18

DT~ (tn.)

Figure 19a

Monotube Pile : C.I.P. Concrete WIth RSA.

87

86

95

94

93

92

91

" ! 00

a- u :t 88 0-

87

ae 86

84

83

82

81 0.05 0.07 0.08 0.11 0..13 O.1ij

aT. (Tn.)

Figure 19b

Monotube Pile: C.I.P. Concrete 35 WIth RSA.

106~----------------"------------------------------~

104

102

100

98

96

94

92

90

88

a8 84

82 80

18

76

14 72 ~4---~----~--~-----1~--~--~1----~---~(----~---~

0.08 om 0.09 0..11 0.13 0.15

01, ell OT. (tn.)

Figure 20

Monotube Pile: C.LP. Concrete wtth RSA.

~,---~-----"-----------------------------~---------------

87

86

86

84-

83

82

81

80

78

78

n;----.----.----r----.I----~--~I----~------~~~ o 20 -+0 eo 80 100

" Skfn Frfctton

Figure 21

PHASE II

Stabilization of a Highway Embankment Using steel Sheetpiling

with steel Batter Piles

I. OB.JECrIVE

To monitor the performance of steel PZ-35 sheeting used in

conjunction with steel HI? 10 x 53 batter piles as a methcxi of

stabilizing a highway embankment against slope movement.

II. BACKGROUND

36

'lhis project developed fran a field condition that was declared

an emergency in the Spring of 1987. '!he site, located on Route 66 in

Portland, COnn., was experiencing recurring highway pavement

settlement which inclinometers installed by Conn ror showed to be

progressive outward slope movements. This phenomenon has been

occurring for the past several decades. Available infonnation

indicated that the greatest slope movements tend to occur during the

spring high-water season, suggesting that pore pressures beneath the

embankment may be a contributing factor. Another contributing factor

may be the dynamic/vibrato:ty loading caused by heavy vehicles passing

the site.

r:Ihe subsurface profile, as shown in Fig. 1, consists mainly of

granular fill overlying peat, which rests on sand. Depth to bedrock

varies fran 20 feet (western edge of site) to 55 feet (eastern

side). Ee.drokc in primairly triassic sandstone of the portland

fonnation. '!he presence of the ol:ganic peat exacerbates the problem.

CDNNror engineers decided in May 1987 to attempt to stabilize

the slope by structural means using steel sheeting supported

~; ,~ / ,_,_ 7;"'" V CQIC C-.L..I Typicel Sec,h'on -Ap?fOX SfC/o 220-1- 10

,_4\=:~£~ '~::. \!~-i

K'1= ~.b'

.... t ('J " ,"

, £fnbcrnt . . "'n'll

5;;.'1 it:. /" -:: 4-() l Plan Treafmenf If-om sta. 219tO -to S-fc-:1. 222+ 75 (h~(¥ox, 375 ~.F.)

---+---1 --.0 . ~~Q. If(

'j(. " II C:;"fJrdFa,"r II --,.---~,:~,~ .. ,~----~------------~-------~

. LTopoP Slope . ~ U

I 22 • ~.fi. -

I . shed pi It. '><fa II viI lip ire PraCC5 6' {);t:.. 7 . :~" .. l II 1.1.' '" 1'1' , I It I ". t f" t t i,'I~::::'='=I='=I_t '_.1-----

c::.wtv/<. Toe. of Slope - . ~ Approx. H(9nwtty Line

Figure 22

38

laterally by steel batter piles. UCONN personnel instnnnented two

sheetpile sections with strain gages to measure the strains induced

in the sheeting. CONNlXYI' mounted slope indicator tubes at two other

locations on the sheeting to monitor the deflection of the sheeting.

UCONN and CONNlXYI' staff inplemented seperate instnnnentation

schemes which, would complement the evaluation of sheetpile

perfonnance.

III. INSTRUMENTATION PROGRAM

a. Underlying '!heo+Y'

'!he inclinometer directly measures vertical slope with respect

to a reference. '!he slopes along the vertical profile can be I

integrated, with an appropriate bounda:ty condition, to give a

deflection curve; or can be nun:terically differentiated using finite

differences to obtain,

Ely" = M, from SlTIa.ll deflection beam theo:ry,

which can be co.n:pared with the strain gage data directly and inplies

that the spatial rate of change of slope is proportional to the

internal bending moment at that section. However, accuracy and

vertical positioning error makes this numerical integration approach

only approximate at best.

b. UCONN Instnnnentation Design and Construction

A strain gage instnnnentation program was developed to detennine

the bending moment distribution along the vertical profile of the

sheeting. Several challenging problems had to be addressed. '!he

instnnnentation had to withstand, a. impact driving by a Vulcan #1

pile driving hammer, b. a subm.eJ:ged envirornnent below the ambient

water table, c. potential damage by vandals.

39

1'--_ .......... ·--9.00 "-_ .... ···· ..... ··--i

_ .... --1 2. 25" 1---4. 25 "~ 2. 25" 1----

L 3*3=:--~-.' ,

1 II

"'1"--,-------.--18.0"-, --- ,---,-----1

Figure 23a

I

i

cP'"

i i 0 I

j

j ~

- .- 0 -+----- ---1 ------ j -e--I \ I

I 1 I ! ! I

o I

I I !

o o

An

. ACl..G ~ fItS) MAulING' &P-N . ~~Lb ~1'{)

-.-. '- A' .

~,,..JG 7LArG7

(JjRCftL'j

Prd-oJa~ Af.>J.m~lY

Figure 23b

Advice was sought from Dr. J. Gartner of the University of

Connecticut Mechanical Engineering Department, a specialist in

instnnnentation, and engineers at Micro-Measurements group of

41

Raleigh, N. C. Electric resistance type strain-gages were selected

for this project. In addition to gage selection and attachment to

the pile, a means for protecting the instnnnentation and its

associated wiring required careful consideration. '!he cover scheme

selected consisted of steel 3" x 3" angles placed longitudinally onto

the flanges of the PZ-35 sheetpiling (Figure 23b). Ten foot long

segments of the angle would be welded to a pile with 1 foot gaps left

between them. within these gaps would be placed the strain gage

arrangment and protective coatings. Following gage installation, a 1

foot piece of steel angle would be bolted to the sheetpiling to

protect the instnnnentation assembly from physical damage during

driving. Any attempt to exclude water from inside the angle sections

would not be successful, hence individual waterproof coatings were

applied to the gages instead.

'!he bonding of the gages to the steel piling was to be

accomplished by spot-welding of the steel strain-gage backing to the

piling, a task which was done in the field. A special layered.

application of sealants was applied to the area surrounding the gages

after the necessaJ:Y electrical connections were made to the strain

gage (see Figure 24). Next, a steel 3X311 angle cover were bolted to

the pile above the gages to help protect the instnnnentation during

pile driving operations.

Data Collection

Soon after pile installation, reference "zerol1 readings were

taken on all strain gages using a Micromeasurements model P-3500

42

Silicone Sealant

Teflon Tape Covering

-_ .. - --- --~ ~~~-=NG __ '-~"'-'--"'''''''--''-"-'''''

Figure 24

digital strain readout unit. Following the next several sets of

observations, it was soon evident that the readings of the gages were

eroneous. '!he problem was that the readout box would not give a

stable reading: the reading drifted. with time during measurement.

'!he possibility of water or moisture getting to and possibly banning

the gages was of primary concern. At this point, Mr. John Fickiet,

Chief Electronic Technician at UCX:>NN, was called in to assist in the

investigation of this problem. Mr. Fickiet concluded that the design

of the box, being of a high ilrpedence input type, was picking up

stray electron:agnetic signals at the location and these signals were

responsible for the drift problem. A low in'pedence type bridge, such

as the older BIH type of wheatstone bridge circuitJ::y was reccommended

as a solution. Measurements taken with the BIH readout box indeed

gave stable strain gage readings and was used exclusively thereafter.

strain gages readings versus pile depth and gage readings versus

elapsed time are shown in Figures 25 thru 32 for both the east and

west instnnnented sheetpiles.

9

8

7

6 . '" c - 5 i 2~ 4 o· _"0 ~c "",0

3 • Ot::J co ijt 2 ~ • 1 Ot 0 0

0

-1

-2

-3 0 20 40

[] Depth - -6

STRAIN VS. TIME East Sheetplle, (Swamp aIde)

60 80 100

Elapsed time In days. Q -26 + -16

Figure .25

120 140

A -36

160

.p.. w

. ,..."..

c 2 .. e" o· _"0 ~c '"'g 0;, co iit f • at

&

STRAIN vs. DEPTH East Sheetplle, (embankment aide)

4 ~------------------------------------------------------------------~

3

2

1

0

-1

-2

-3

-4~1~------~--~--~--~--T---~--~--~--~--~--~---r--~--~

-~ -32 -28 -24 -20 -16 -12 -8

Elevation wlth Respect to Grade C 23 + 37 <> 93 A 120 X 151 Days

Figure 26

.j>­

.j>-

.. " c -e 1i e" o· _"0 ;:ec '"'i 'O:s co ;;,t, f • 0

8

STRAIN vs. TIME East Sheetplle. (embankment aide)

4~i------------------------------------------~

3

2

1

0

-1

-2

-3

-4 o 20 40 60 80 100 120 140 160

C Depth = -6 Elapsed time In days.

+ -16 ~ -26 A -36

Figure 27

..p.. U1

STRAIN vs. TIME 2

west Sheetplle. (Swamp _Ide)

1

0

• '" -1 c -2 11 e~ -2 S" 2C ,-,0

-3 • Ut:J co It -4 ! • CIt -5 0 ()

-6

-7

-8 0 20 40 60 80 100 120 140 160

C Depth = 6 Elapsed time In daye.

+ -4 ~ -14 A. -24 .p-O'.

Figure 28

1.3

1.2

1.1

1

• 0.9 '" c -2 0.8 1» 2~ .2"0 0.7 ::EC '-'0 0.6 o! co 'Db 0.5 e

0.4 • OIl 0 0.3 0

0.2

0.1

0

-0.1 -24

[] 21

STRAIN VS. DEPTH west Sheetplle, (embankment .fde)

-20 -16 -12 -8 -4

Elevation with respect to grade + 37 <> 93 A 120

Figure 29

o

X 151 days

4

.+:'­-.....J

2

1

0

. " -1 c -e 1i e" -2 o· _'0 ~c ,-,0

-3 • Ot:J co ;it -4 !

• 0' -5 0

" -6

-7

-8

-24

·0 21

STRAIN VS. DEPTH west Sheetptle. (Swamp side)

-20 -16 -12 -8 -4

Elevation with Respect to Grade + 37 ~ 93 A 120

Figure 30

o

X 151 Days

4

+:-­co

9

8

7

6 . " c -e 5 ..-2" 4 o· _"0 ~c .....,0

3 • "':3 co ~t 2 ! • 1 Qt 0 ~

0

-1

-2

-3 -36

C 21

STRAIN vs. DEPTH East Sheetplle. (Swamp aide)

-32 -28 -24 -20 -16

Elevation with Reepect to Grade + 37 <> 93 It. 120

Figure 30

-12

X 151 Days

-8

+:-­~

. 1.3

1.2

1.1

1 .

0.9 " c -f 0.8 • e~ 0.7 .2"0 :i C

- ,-,0 0.6 • 0:1 co

:gt 0.5 !

0.4 • 0 a 0.3 C

0.2

0.1

0

-0.1

STRAIN VS. TIME west Sheetpn •• (embankment aide)

0 20 40 60 80 100

Elapsed time In days. C Depth = 6 + -4 ~ -14

Figure 31

120 140

A -24

160

\J1 o

51

calculations indicate that the maximum internal bending moment

developed in the PZ-35 sheetpiling occurs at an elevation

approximately 10 feet above the pile tip. In addition, the internal

moment at the pile tip and pile butt appears to be quite small

suggesting that conditions at these points approach those of pinned

connections. Observation of the strain gage reading versus time

plots (Figs. 25 thru 32) suggest that the pile-soil system, and slope

has stabilized. 'Ihis is further substantiated by obse:r:ving that

vertical settlement of the Route 66 pavement surface appears to have

been arrested.

Bending moments were calculated by using linear-elastic

assu:rcptions together with simple beam theo:ty; with

M = S E 8b~ where 8b = (8 S - 8 e ) / 2

8 S = strain on swamp side of sheeting

8 e = strain on embankment side of sheeting

81;) = strain due to bending

E = Young's modulus

S = section modulus of sheeting (per foot of wall)

M = internal moment in sheetpile

Independent inclinometer readings support the general shape of

the moment versus depth cu:tVes. However; the magnitude of the

bending moments do not agree with infonnation obtained from

inclinometer readings taken by eonnooI' personnel.

'!he measured bending moments obtained from the strain gages are

several times higher than those back-calculated fran the

inclinomenter readings. Water infiltration onto the strain gages and

associated electrical connections was investigated as a possible

source of error. In light of the harsh envirornnental conditions

52

present, moisture may have worked its way onto the strain gages although

the methcxl of waterproofing recommended by the manufacture was used.

'!he university of COnnectiOlt geotechnical engineering laboratories

devised a simple laboratory experiment to measure the effect of moisture

upon gage performance. '!he perfonnance of a strain gage was measured by

inunersing the gage and associated electrical connections in a brackish

water bath (to model in-situ corrlitions) • '!he results, when conpared to

readings taken while the gage was dl:y, indicated that the microstrain

readings were essentially the Same after allowing for an initial drift

pericxl of several minutes duration for the gage to stabilize.

Electrically, water acts as a parallel shunt resistance across the 120

ohm strain gage grid which makes the effective resistance of the gage

slightly less than 120 ohms. Theoretically, this shunting effect should

lessen the sensitivity (gage factor) of the strain gage, and apparent

readings should be less than those actually occurring. However, as

detennined from the laboratory experiments, this effect is quite small.

Therefore, water seepage can be tentatively ruled out as the primary

reason for the abnonnally high strain gage readings.

Conclusions

1. The maximum bending moment in the sheetpile appears occur about 10

feet from the bottom of the pile.

2. The magnitude of the maximum moment is uncertain as strain gage

measurements and inclinometer data do not compare well.

3. '!he batter/sheetpile wall appears to have stabilized the slope

against continued outward movement.

4. The Continued use of electrical resistance type strain gages in

harsh environments is not reconunended.

REFERENCES

1. Smith, E.A.L. (1960), "Pile Driving Analysis by the Wave Equation", J .S.M.F.D., A.S.C.E., Vol.86, SM4: pp. 35-61.

2. Kolsky, H. (1963), STRESS WAVES m SOLIDS, Dover, New York.

3. Pochhammer, L. (1876), "J. Reine Angew" Math, Vol. 81, P. 324.

4. CAIWAP - Computer Analysis Program

5. Forehand, P.W. and Reese, J .L., (1964), "Pred.iction of Pile capacity by the Wave Equation", J.S.M.F.D., A.S.C.E. Vol. 90, 8M2, pp. 1-25.

6. Peck, R., Hansen, W. and Thombum, T., (1974), FOUNDATION ENGINEERING, 2nd ed., New York, Wiley, p. 215.

7. Van der Veen, C., (1953), "The Bearing capacity of a Pile", Proceedings of the Third Int. Conf. Soil Mech. Fed. Eng., Vol. 2, pp. 84-90.

8. Terzaghi, K., and Peck, R.B., (1967), Soil Mechanics in Engineering Practice, 2nd ed., Wiley, New York.

9. O'Neill, M., Unpublished information.

··53

10. Holloway, D.M., Clough, G.W., and vesic, A.S., "The Mechanics of Pile-Soil Interaction in Cohesionless Soils", Report for the Co:rp. of Engineers, Vicksburg, Miss., Dec., 1975.

REFERENCES

1. Smith, E.A.L. (1960), "Pile Driving Analysis by the Wave Equation", J.S.M.F.D., A.S.C.E., Vol.86, SM4: pp. 35-61.

2. Kolsky, H. (1963), S'rnESS WAVES rn SOLIIlS, D:wer, New York.

3. Pochh.a:rraner, L. (1876), "J. Reine Angew" Math, Vol. 81, P. 324.

4. c.ru::wAP - computer Analysis Program

5. Forehand, P.W. and Reese, J.L., (1964), "Prediction of Pile Capacity by the Wave Equation", J.S.M.F.D., A.S.C.E. Vol. 90, 8M2, pp. 1-25.

6. Peck, R., Hansen, W. and'Ihornbmn, T., (1974), FOUNDATION ENGINEERING, 2nd ed., New York, Wiley, p. 215.

7. Van der Veen, C., (1953), "'!he Bearing Capacity of a Pile", Proceedings of the Third Int. Conf. Soil Mech. Fed. Eng., Vol. 2, pp. 84-90.

8. Terzaghi, K., and Peck, R.B., (1967), Soil Mechanics in . Engineering Practice, 2nd ed., Wiley, New York.

9. O'Neill, M., Unpublished information.

10. Holloway, D.M., Clough, G.W., and Vesic, A.S., "'!he Mechanics of Pile-Soil Interaction in Cohesionless Soils",.' Report for the Corp. of Engineers, Vicksburg, Miss., Dec., 1975.