experimental use of high-strength reinforcing...

Experimental Use of High-Strength Reinforcing Steel

by

Jere G. Newton formerly

Associate Designing Engineer

and

Larry G. Walker Supervising Design Engineer

Research Report Number 25-1 F

Experimental Use of High-Strength Reinforcing Steel Research Project 5-62-25

Conducted by

Bridge Division The Texas Highway Department

In Cooperation with the U. S. Department of Transportation

Federal Highway Administration Bureau of Public Roads

May 1967

The opinions, findings, and conclusions

expressed in this publication are those

of the authors and not necessarily those

of the Bureau of Public Roads.

ACKNOWLEDGEMENTS

The design and test of phases of this experimental

project were supervised by Mr. E. L. Hardeman, Senior Bridge

Engineer for District 9. Mr. W. R. Casbeer designed the ex

perimental structure and Mr. J. N. Newton was responsible for

field tests and data analysis. The research reported was

performed in cooperation with the Bureau of Public Roads.

The valuable assistance and cooperation of the Bureau

of Public Roads in planning and executing the field tests is

greatly appreciated, particularly the work of Mr. R. F. Varney

and Mr. C. F. Galambos who were in charge of live load

testing and equipment.

TABLE OF CONTENTS

LIST OF FIGURES AND TABLES • • • • • • · . ABSTRACT • • • • • • • • • • • • • • • • · . . . . I.

II.

INTRODUCTION • . . . . . . . . . . . . . . . . DETAILS OF THE EXPERIMENTAL STRUCTURE · . Description •••••••••• • • • • • Material . . . . . . . . . . . . . . . . . Des ign • • . .. . . • • • • • . . . . . . Construction • • • • • • • • • • • • •

III. FIELD TESTS AND EVALUATIONS · . . . . . .

IV.

Instrumentation •••••••••• Dynamic Live Load Tests •••••

. . . Dead Load and Time Dependent Deflections • Crack Width Measurements •••• • • •

• • · .

DISCUSSION OF RESULTS Dynamic Live Load Tests Dead Load and Time Dependent Crack Width Measurements • •

· . . . . . . · . . . . . . . . Deflections • · . . . . . . . .

V. CONCLUSIONS . . . . . . . . . . . . . . . . . APPENDIX • • • • . . . . .

References • • • • • • • • • • Figures and Tables • • • • • •

• • • · . . . . • • • · . . . .

Page No. ii

v

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3 3 4 4 6

7 7

10 13 13

15 15 17 18

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Figure No.

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2

3

4

5

6

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8

9

10

11

12

13

14

15

16

17

18

19

20

21

LIST OF FIGURES AND TABLES

Structure Layout

Structure Details

Structure Details

Structure Details

Strain Gage Installation

Deflection Gage

Axle Loads and Wheel Spacing

Stress Distribution

Stress Distribution

Stress Distribution

Stress Distribution

Deflection Distribution








Peak Deflection vs. Truck Speed


ii

Page No.

27

28

29

30

31

31

32

35

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42

43

44

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46

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52

53

Figure No.

22

23

24

25

26

27

28

29

30

31






Observed Deflection Due To Dead Load

Observed Deflection Due To Dead Load

Deviation From Predicted Crack Width vs. Occurence Frequency

Crack Width vs. Occurence Frequency For Each Bent (Before Traffic)

Crack width vs. Occurence Frequency For Eacy Bent (After Traffic)

Observed Crack Width vs. Occurence Frequency For Each Bent (Final Reading)

iii

Page No.

54

55

56

57

58

59

61

62

63

64

Table No.

1

2

3

4

5

6

7

8


Steel Stress (PSI) In Bottom Of Girders Center of Span 2

Steel Stress (PSI) In Bottom Of Girders Center Of Span 4

Deflection Of Center Girder (Inches)

Observed Impact Factors

Observed Impact Factors

Vibration Frequency Of Test Vehicle During Test Runs (CPS)

Peak Double Amplitudes Of Vibration Measured On Truck Axles

Forced Vibration Frequency Of Bridge (CPS)

iv

Page No.

33

34

39

48

49

50

51

60

A B S T R ACT

The design, construction and study of a full scale ex

perimental overpass structure is reported.

A-432 reinforcing steel having a minimum yield strength

of 60,000 psi was used in the four span (55'-88'-88'-55')

continuous haunched concrete girder unit. Ultimate strength

theory was used in the desig~and stresses were also checked

by elastic methods.

Electrical strain gages were installed at selected

points on the reinforcing steel and deflection gages em

ploying electrical strain gages were mounted on the com

pleted structure. Oscillographic re~ords of strain and de

flection were taken for a series of live load tests.

A three axle truck was used in the live load tests.

Speed and lane position were varied for purposes of studying

impact, stress distribution, vibration characteristics and

deflection characteristics. Comparisons of observed and cal

culated stresses and deflections were made.

A study of crack width and distribution was made at

several times during the test program. Observed crack forma

tion is compared with various crack prediction methods.

Time dependant deflections were observed throughout the

test period and compared with deflection predictions.

v

The performance of the structure appears to be satis

factory. Crack formation is not considered excessive and

observed strains indicate an adequate factor of safety against

static failure.

vi

Report On

EXPERIMENTAL USE OF HIGH STRENGTH REINFORCING STEEL

I. INTRODUCTION

The successful use of high strength reinforcing steels

in Europe and in building construction in this country

prompted the Portland Cement Association and other agencies

to promote its use in bridges. Laboratory tests have indi-

cated that the high strength steels manufactured in this

country would perform satisfactorily under bridge loadings.

In an effort to gain experience in design and construction

with high strength steel and to evaluate its performance, the

Texas Highway Department initiated plans for a full scale

experimental structure. The crossroad structure over Inter-

state Highway 35 near Hillsboro, Texas,was built as a part

of the Interstate Highway Program and the experimental

phases of the project were accomplished in a research project

in cooperation with the Bureau of Public Roads.

Four structural requirements that must be satisfied by

all reinforced concrete bridges have been listed by Fountain

in an unpublished paper (Reference 1). These are:

1. Adequate safety against static failure

2. Adequate safety against fatigue failure

3. Satisfactory crack formation

4. Satisfactory deflections

It is the broad objective of the research reported to

evaluate the performance of the experimental structure with

respect to these requirements. However, no specific tests

were directed toward evaluation of safety against fatigue

failure.

2

3

I I. DETAILS OF THE EXPERIMENTAL STRUCTURE

Description

A four span continuous concrete T-Beam unit having span

lengths of 55 1 -88 1 -88 1 -55 1 was chosen for this study. The

6~" thick slab was cast monolithically with the parabolic

haunched girders. Three 21_4~" wide girders supported the 241

roadway. The total girder depth was 2 1 -3 11 in the uniform

depth sections and increased to 4 1-9" over supports. The

design loading was H15 for this 30 degree skew structuIE. A

general layout of the structure is shown in Figure 1 and

structural details are shown in Figures 2, 3 and 4.

The interior supports consisted of a single 30" round

column under each girder. Hinge action between columns and

girders was achieved by placing neoprene cushions on the outer

edges of the columns and by placing connecting dowels perpen

dicular to girders at the center of columns. The U-type

abutments and the 3~'' columns \lSre supported on drilled shafts

(cast-in-place piles) •

Several structures of this type and of conventional de

sign and reinforcing were constructed on the same project.

One of these, a four span continuous concrete T-Beam unit

having spans of 55 1-761_76'-55 1, was chosen as a control

structure for comparison purposes.

4

Material

High Strength A432 reinforcing steel with a minimum

yield point of 60,000 psi was chosen for the main steel in

the girders and deck slab. Regular strength A15 reinforcing

was used for stirrups and for column reinforcing. Concrete

with a minimum compressive strength of 3,000 psi and a mini

mum cement content of 5 sacks per cubic yard was specified.

Design

The continuous girder unit was designed by ultimate

strength theory. Load factors of 1.5 for dead load and 2.5

for live load were used except in the cases where dead load

and live load were of opposite sign. In these cases the

dead load factor used was 0.5. Moment distribution was based

on gross section moment of inertia. Lateral distribution of

wheel loads followed AASHO design criteria.

Comparison designs were made by elastic theory for both

30,000 psi and 20,000 psi working stress for reinforcing

steel. Tension steel requirements determined by ultimate

strength theory compared closely with those determined by the

elastic theory with 30,000 psi working stress. The major dif

ference in reinforcing steel requirements was in compression

reinforcement over supports. No compression steel was re

quired by ultimate theory; whereas, 14.62 sq. in. was required

5

by elastic theory. The conventional percentage of positive

moment steel was extended into supports, however. Bars in

girders were cut off as indicated by requirements and no bars

were bent up.

Ultimate strength theory was used for design of the deck

slab but shear stirrups were designed by conventional elastic

design methods.

Top girder steel was distributed over the full width of

girder flanges. Three dif nt bar sizes were used to satis-

fy the tension steel requirements over supports (#7 over the

first interior support, #8 over the center support and #9

over the third). This was done to evaluate the effect of

bar size on crack width. Crack width predictions were com

puted by several empirical formulas for comparison with ob

servations.

Deflection predictions based on gross moment of inertia

were compared with predictions based on transformed moments

of inertia. Both methods gave quite similar deflection pat

terns. The value of modulus of elasticity normally used in

determining construction camber for this type structure is

1.5xl06 • This value was intuitively decreased to 1.Oxl06

for this structure effectively increasing construction cam

ber.

construction

All forming and falsework for the superstructure was

completed prior to beginning concrete placement. All super

structure concrete was completed in seven placements during

the period June l8-June 28, 1963, and all forms were re

leased on July 8, 1963. The sequence of placement is shown

on Figure 2. Deck placements were finished by longitudinal

screeding. The average 28-day compressive concrete strength

was greater than 5,000 psi.

6

III. FIELD TESTS AND EVALUATIONS

Instrumentation

Electrical Strain Gages were bonded to selected rein

forcing bars prior to their placement in the forms. The

gages used in this test were Baldwin-Lima Hamilton A5-1S6,

FAP-50-12-S6, AS-9 and ES-9S gages. The A-5 and FAP-50

gages were temperature compensated for steel; the ES-9S

gages were temperature compensated for concrete. The gages

were mounted on bars ranging from #4 through #11 with the

ES-9S gages placed directly in concrete.

7

The bars were prepared by grinding the deformations off

of half the circumference of the bar for a length of four to

six inches. This area was then filed to remove surface

irregularities. Acetone was used to remove dirt and grease.

An epoxy and adhesive, formulated by the Texas Highway

Department Laboratories, was used to bond the gages to the

bars. This was a two component epoxy which was proportioned

on precise laboratory scales and stored in separate eight

ounce cups. The epoxy could then be field mixed with the

critical proportions needed for complete hardening.

Care was exercised in the handling of the gages to avoid

getting oil from the hands on the gages and breaking the

bond. The gage was glued to the bar and worked into place

carefully to avoid air pockets in the epoxy. The epoxy was

allowed to cure for 24 to 48 hours.

Four wire, 25 gauge cable was then soldered to the

8

gages using two leads to each gage wire. The ground wire

from the cable wag soldered to the reinforcing bar to facili

tate resistance to ground checks.

Di-jell wax was used to waterproof the gage and the

soldered joint. This wax was melted and applied by drops to

avoid air pockets.

The cables were epoxied to the reinforcing steel so that

forces applied to the cables would not be transmitted to the

delicate gage leads. A two component flexible epoxy, also

formulated by the Texas Highway Department Laboratories, was

applied over the area to protect the gage and soldered con

nection from mechanical damage.

Figure 5 shows the method of installation and protection

of gages. This method of installation did not prove entirely

satisfactory in protecting gages from mechanical damage

during steel and concrete placement. A number of gages were

damaged beyond repair. Also, the majority of the ES-9S gages

placed in the concrete were destroyed.

The installation of the strain gages was completed

several months before the actual construction of the

9

superstructure began. The gages were checked prior to the

beginning of construction and all bad gages were repaired or

replaced.

Lead cables were brought out of the superstructure at

the center column of each of the three interior bents. The

lead cables were taped to the reinforcing bars to minimize

possible damage from pouring and vibration.

Dummy or temperature compensating gages were mounted on

6" sections of reinforcing bars in the same manner as the

active gages. Extra coatings of flexible epoxy were applied

to the ends of these sections so that no stress would be

introduced through end bearing. These gages were then placed

as close as possible to their respective active gages so that

the temperature effect would be the same for both gages and

both lead wires.

The cables were brought down the columns into terminal

boxes strapped to the columns. The cables were soldered to

banana jacks mounted on a sheet of plexiglass. The ground

wires were shunted together to facilitate resistance to

ground tests.

Dynamic deflection measurements were accomplished with

devices consisting of cantilever arms clamped to the struc

ture and initially deflected by a wire attached to a fixed

10

ground reference {Figure 6} • Four strain gages were

attached to the arm to sense changes in deflection of the

arm. These assemblies were then calibrated to indicate

structure deflections. It was necessary to correct for

changes in the stress on the reference wire due to structure

movement.

Equipment for dynamic loading and for recording dynamic

strains and deflections was furnished and staffed by the

Bureau of Public Roads. Loading was accomplished with a

three axle truck which was equipped with instrumentation for

measuring dynamic axle loads. The physical dimensions of

the truck are shown in Figure 7. It could be loaded to ap

proximate either H15-S12 or H20-Sl6 loading.

The strain recording equipment had a capacity to

monitor over 50 strain gages and/or deflection indicators

simultaneously. Gage output was amplified to drive light

beam galvanometers which produced oscillographic records on

light sensitive paper. It was possible to adjust the ampli

fication and thereby control the amplitude of trace move

ments on the paper. This instrumentation was calibrated

periodically.

Dynamic Live Load Tests

The two-lane structure was divided into five test lanes.

Lanes 1 and 5 were along the east and west curbs re

spectively, with the tires of the truck approximately 8"

from the curbs. Lanes 2 and 4 were the regular traffic

lanes and Lane 3 was along the centerline of the structure

(Figure 1) .

The Bureau of Public Roads test truck made 620 test

passes over the five test lanes, each pass being called a

run. During these tests, six different sets of gages were

monitored. The runs for each set are referred to as a

series.

11

The loading for each series was constant, either H15-S12

or H20-S16. The speed and lanes were varied in each series.

Lanes 1 and 5 were used only for creep speed runs because of

their nearness to the curbs.

Test loading was done before the bridge approaches were

completed. Test runs were made downhill from north to south

over a temporary gravel fill, which appeared to cause exces_

sive bouncing of the truck. Severe impact was induced on 27

of the H20-S16 runs by placing ramps in the wheel paths

which elevated each axle and allowed it to drop 2" abruptly.

The impact ramps were alternately placed at the centerline

of the two middle spans, at the 0.6 point of the first span

and over each of the interior bents. Only one location was

used during anyone run.

12

Three air hoses were mounted perpendicular to the road

way at the beginning, center and end of the bridge to record

truck location on the oscillograph records. An event mark

was entered on the record as each axle passed over each hose.

These air hoses also actuated a digital time counter which

timed the passage of the truck to the nearest 1/1000 of a

second.

The lateral displacement of each run from the intended

test lane was measured by placing a series of pins at the be

ginning and end of the intended lane. An arm hanging from

the center of the truck tripped one of the pins as the truck

passed indicating lateral position of the truck.

Manual observations that were recorded for each run in

cluded time of day, truck speed, test lane, lateral displace

ment, and calibration data. Digital data reduction equipment

was used to transfer selected oscillographic trace readings

to punch cards. The oscillographic record rolls could be

mounted on the equipment and easily manipulated with the vari

able speed two directional chart drive. The setting of

vertical and horizontal indicators on a desired trace point

produced two analog outputs which were automatically con

verted to digital output and punched on cards along with

certain preset identification data. A computer program was

13

written to scale and convert digital trace readings to

stresses or deflections. The program output of stress or de

flection included tabulations by test runs and by individual

gages.

Dead Load and Time Dependent Deflections

Accurate levels were taken on the deck surface before

and after form removal and periodically thereafter to deter

mine initial dead load deflections and time dependent de

flections. The tenth points in each span were marked off

along five lines: the centerline, one foot from each curb

and the top of curb lines. Levels were taken at the same

points on each occasion.

Crack Width Measurements

Crack measurement was accomplished by use of a magni

fying comparator. This device is a microscopic viewer in

which lines of known width can be compared with cracks to

determine widths. By this method the crack width could be

measured to the nearest .001 inch at the point of measure

ment. Crack measurements were taken only at points of maxi

mum width. No measurements were made at points where edge

spalling of the crack had occurred since this was not con

sidered representative of the crack width below the surface.

The cracks were first marked by tracing each crack with

14

a felt marker. This trace followed the entire length of the

crack and was drawn as close to the crack as possible. This

permitted quick location and identification of each crack

and also enabled the crack pattern to be photographed.

Points of maximum-width were then located and marked so that

the same point would be measured each time. Measurement of

crack lengths was also made and a terminal reference mark

used so that crack extension would be readily visible.

The number system used to identify the cracks was sub

scripted so that primary, secondary and tertiary cracks could

be distinguished. primary cracks were marked before traffic

was placed on the bridge but after form release.

Secondary cracks were identified after a series of

"shake down" runs with an H1S and H20 truck. This included

some runs over the 2" impact ramps. Cracks were then marked

and measured at intervals during the test and after the last

run had been made.

Cracks were marked and measured on both top and bottom

of the slab in the negative moment area and on the bottom of

the girder in the positive moment area.

15

IV. DISCUSSION OF RESULTS

Dynamic Live Load Tests

Reinforcing Steel Stress. Significant results of live

load stress measurements are summarized in Tables 1 and 2.

Typical lateral distribution of stress level is shown in

Figures 8 through 11. The combination of stresses

(Figure 11) produced by test runs in lanes 2 and 4 most

nearly approximates design loading assumptions.

Maximum stresses occurred when the H20-S16 test truck

passed over the impact ramps. Even in this case, recorded

stresses combined with computed dead load stresses were less

than calculated values and less than the 30,000 psi value as

sumed for elastic design comparison. It should be noted

that the skew effect and the high concrete strengths would

produce a stiffer structure and that the design assumption

of cracked section was undoubtedly not present at all gages.

Deflections. Typical live load deflection measurements

are shown in Table 3. The effect of truck position on the

peak deflection of each girder is presented in Figures 12

through 17. The computed values shown on these figures were

determined by calculating the deflection due to one test

truck on one girder and distributing this effect to the

three girders in the average proportion of measured de

flections.

The accumulated effect of runs in Lanes 2 and 4 may be

seen in Figures 18 and 19. The skew effect and increased

concrete strength and modulus of elasticity (Ec observed =

5.5 x 106 ) contribute to reduction of deflections.

Impact and Vibration Characteristics. Measured impact

as indicated in Tables 4 and 5 was in general agreement

16

with values computed by A.A.S.H.O. design criteria although

it was highly inconsistent. This inconsistency can be

partly attributed to the oscillations of the test truck as

it came onto the structure. As was mentioned previously,

the approaches had not been completed at the time of testing

and a temporary gravel fill was used. The vibration fre

quency and peak double amplitude of vibration measured on

truck axles are presented in Tables 6 and 7.

The impact measurements in Lane 2 were consistently

high, which may be due to a rougher riding surface in that

area. There was no consistent trend of increased impact

with increase of speed. Speed had very little effect on

peak deflection, as may be seen in Figures 20 through 25.

No unusual vibration characteristics were revealed by

these tests. The forced frequency of vibration of the

structure ranged from 2.50 cps to 3.89 cps, as shown in

Table 8. The free vibration frequency was about 3.5 cps

for the interior spans.

The damping characteristics of the experimental struc-

17

ture were good. A log decrement of damping ranging from

.109 to .147 was observed. These values should be con

sidered as approximate since very low amplitudes were being

measured. The values of log decrement of damping given

above correspond to percentages of critical damping of 1.73%

i...;.td 2.33% respectively. This compares to a range of 1.4%

to 2.0% observed on the A.A.S.H.O. Road Test (Reference 2)

for conventionally reinforced concrete bridges.

Dead Load and Time Dependent Deflections

Deflections observed approximately two years after con

struction were less than the calculated values used in

setting construction camber. The calculated values were in

tuitively increased as described previously to adjust for re

duced reinforcing steel area. The adjusted calculation re

sulted in a predicted final interior span deflection of

0.234 ft., as compared to an observed value of 0.158 ft.

after two years. The observed increase in deflection from

April 1964 to June 1965 was 0.047 ft. Observed deflections

are presented in Figures 26 and 27.

18

As in live load tests, high concrete strengths and high

modul s of elasticity contributed to reduced deflections,

but it would appear that the effect of reduced steel per-

centage was overestimated in design calculations.

Crack width Measu~ements

The crack width study reported here is limited to the

negative moment area cracks only. The positive moment area

measurements were considered unreliable since the girders

were rubbed shortly after form removal. The cracks that did

reappear were not considered to be representative of the

actual crack pattern.

Observed crack widths were compared with theoretical

widths predicted by four crack width formulas. These formu-

las are given below:

(1) Watstein and Parsons with constants evaluated by

Clark (Reference 3)

W= 2.271110-8 h~d x p [fs-S6.6(i+ n >]

(2) Revised (1959) European Concrete Committee formula-

CEB (Reference 4)

(3) Jonsson, Osterman and Wastlund formula (for pure

bonding) 1 f 2/3

Wmox = KxD [*' xfs"]

(4) Kaar and Mattock formula (Reference 6)

W = 0,077 ~ fs x 10-6

Where: W = Average crack width (inches)

Wmax = Maximum crack width (inches)

h = OVerall depth (inches)

19

d = Compressive face to tensile steel (inches)

D = Bar diameter (inches)

p = As

Ac

fs = calculated steel stress (psi)

n = Es

p e

~

= As Ac (effective)

AC (effective) is the concrete surrounding the tensile bars having the same centroid as the tensile bars (usually = 2b x cover)

6 K2 = 47.5 x 10 for deformed bars

K = 0.23 for plain bars

K = .016 for deformed bars

Ic = Gross concrete moment of Inertia (in 4)

et = Distance from the neutral axis to the extreme fiber of the tension face (inches)

4D2 A = :r:rPe

20

The formulas of Watstein and Kaar are for average crack

width of the primary cracks, while the other two equations

are for the maximum crack width at any point. A factor of

1.5 was used to relate maximum to average values for purposes

of comparing the four equations. This factor was observed

by Hognestad for steel stresses of 30 to 40 Ksi (Refer-

ence 5). The dead load steel stresses here were not in this

range but no other data was found to correlate average and

maximum crack widths of lower steel stresses.

None of the formulas used gave a consistently reliable

picture of the average crack width. The formula of the

European Concrete Committee yielded the best values as it

consistently predicted wider crack widths than the other

formulas (Figures 29 through 31) •

The maximum crack widths were also closer to the CEB

formula. All formulas, however, gave low predictions of

maximum width. This is partly due to the presence of a live

load in the loading history of the bridge. None of the

formulas provide for the existence of short duration live

loads. However, even before the live load was applied, the

maximum theoretical widths were exceeded by 20 to 60 percent

of the observed cracks.

As can be seen from Figure 28, the formula of the CEB

21

yielded the lowest error. On the other hand, the formulas

of Watstein and Kaar have the lowest standard deviation of

error with .00122 and .00124 respectively. This compares

to the .00134 and .00138 in standard deviation of errors by

CEB and Jonssons· formulas.

Although attempts to predict crack width were not sat-

isfactory, the crack pattern and crack widths observed on

the structure indicate satisfactory performance. The maxi-

mum crack width observed after two years was .010 in. Only

two observations exceeded the European Concrete Committee

of .008 in. The average width after two years was .0055 in.

Average primary crack widths observed at various times are

tabulated below:

Date August, 1963 September, 1963 April, 1964 June, 1965

Bent 2 (#7 Bars)

.0037

.0041

.0031

.0062

Bent 3 (#9 Bars)

.0033

.0038

.0032

.0053

Bent 4 (#8 Bars)

.0046

.0046

.0025

.0049

The measurements in August 1963 were made before any traffic

had been placed on the structure. The measurements in

September 1963 were made a r all live load testing was

completed and the structure was opened to regular traffic.

As may be seen from the tabulation of crack widths a-

bove, the use of three different bar sizes did not show a

consistent trend as to their ability to control cracking.

22

Both the August and September 1963 readings yielded smaller

average crack widths in the area reinforced with #9 bars.

The last set of readings shows the largest crack width to be

in the area reinforced with #7 bars.

This is contrary to the performance demonstrated in a

number of laboratory tests which indicate that smaller bars

are superior in controlling crack widths. The final average

crack spacing for primary cracks was 1 1-6", 11-8" and 2'-0"

for Bents 2, 3 and 4 respectively. This is also contrary to

expectations.

Contributing factors to these discrepancies may include

variation in concrete quality and weather conditions, the

sequence of form release and the fact that most of the im

pact runs during live load testing were made on the first

two spans. It will also be noticed that the dead load de

flection of span 2 is greater than that in span 3 (Figure 2~.

The fact that forms were left in place longer than usual be

cause of unrelated construction scheduling factors may have

influenced crack formation. The concrete strengths were

high at the time of form release and this may have helped

to reduce crack width.

Observations of the control structure with regular rein

forcing steel revealed crack formation quite similar to that

on the test structure.

v. CONCLUSIONS

Based on the observations and data collected in this

study of a full scale experimental bridge constructed with

A-432 reinforcing steel, the following is concluded:

23

1. Measured live load stresses under overload conditions

indicate satisfactory working load performance. The

H15 designed structure was subjected to H20-S16

loading over impact ramps without causing excessive

stress.

2. Dead and live load deflections were smaller than

design predictions. This study indicates that the

reduced steel percentage resulting from the use of

high strength steel has little effect on deflections

in the working load range and that conventional

methods of calculating deflection are adequate.

3. Measured impact percentages were generally less than

A.A.S.H.O. design specification requirements.

4. Damping characteristics of the structure were good

and no unusual vibration characteristics were ob

served.

5. Observed crack pattern and widths indicate satis

factory performance.

6. Nothing in the observed behavior of this structure

24

, indicates unsatisfactory performance.

7. Cost experience on this project indicates consider-

able cost advantage in the use of high strength

steel in this type structure. An estimated 30 per-

cent reduction in reinforcing steel resulted. Some

reduction in concrete quantity was also possible due

to being able to reduce width of girder stems. The

contractor's unit price bid for A-432 steel was 9.8

cents per pound in place, which was the same as the

price for A15 steel. The cost of the structure in

place (excluding cost of drilled shafts, riprap,

railing and armour joints) was $4.32 per sq. ft.

The overall saving for the structure was estimated

at 10 percent.

APPENDIX

References

Figures and Tables

25

REFERENCES

1. Fountain, R. S. "High Strength Reinforcing Steel For Bridges," (Unpublished paper) •

26

2'. "Dynamic Studies of Bridges on the A.A.S.H.O. Road Test," Highway Research Board Special Report 71.

3. Clark, Arthur P. "Cracking in Reinforced Concrete Flexural Members." ACI Journal Vol. 27, No.8 (April 1956).

4. Gaston, J. R. and Hognestad, Eivind. "High Strength Bars As Concrete Reinforcement Part 3. Tests of Full Scale Roof Girder," PCA Journal Vol. 4, No.2, 10-23 (May 1962).

5. Hognestad, Eivind. IIHigh Strength Bars As Concrete Reinforcement, Part 2. Control of Flexural Cracking," Journal PCA Vol. 4, No.1, 46-63 (January 1962) •

6. Kaar, P. H. and Mattock, A. H. "High Strength Bars As Concrete Reinforcement, Part 4. Control of Cracking." Journal PCA, Vol. 5, No.1, 15-38 (January 1963) •

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48

TABLE 4

OBSERVED IMPACT FACTORS

Based on Reinforcing Steel Stresses

Impact Factor * Impact

Gage Location ,-,ane # 10 MPH 15 MPH 20 MPH 25 MPH Boards Remarks Bottom of Girder Center of Span 3 2 1.00 1.02 1.00 1.19 - H20-S16 Loadings

It 3 .98 - .97 -It 4 .95 1.00 .99 1.22 -

Top of Slab over Bent 3 2 - 1.40 - - -

" 3 - 1.12 - -II 4 - 1.09 - - -

Top of Slab over Bent 4 2 1.26 - - -

II 3 .91 - - - -" 4 1.19 - - - -

Bottom of Girder 4/10 point Span 4 2 1.10 1.20 1.29 1.17 -

II 3 .98 .95 .93 .83 -" 4 1.03 .91 .89 .89 -

Top of Slab 1.13 1.19 1.86

Impact Boards over Bent 2 2 1.00 11.16 Center _.f.' ...

II 3 1.09 1.15 1.27 1.03 1. 78 " II 4 1.03 1.00 1.06 .97 1.66 "

*Impact Factor is defined as the ratio of stress at the speed indicated to the stress at crawl speed.

at -

experimental use of high-strength reinforcing...

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