Engineering Structures 32 (2010) 1814–1820

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Engineering Structures

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Case study: Damage of an RC building after a landslide—inspection, analysisand retrofitting

P. Tiago a,b, E. Júlio c,d,∗a EC+A – Projectos Lda, Coimbra, Portugalb Civil Engineering Department, Polytechnic Institute of Coimbra, Portugalc ISISE – Institute for Sustainability and Innovation in Structural Engineering, Portugald Civil Engineering Department, University of Coimbra, Portugal

a r t i c l e i n f o

Article history:Available online 5 March 2010

Keywords:DamageBuildingLandslideInspectionRetrofittingRobustness

a b s t r a c t

In 2000, due to a substantial landslide, the reinforced concrete (RC) structure of a residential buildinglocated in Coimbra, Portugal,was severely damaged. The first two levels of three columnswere completelydestroyed and, as a result, part of the building supported by these, with a dimension in plant of 9.5 ×6.7 m2, became a 7.0 m span cantilever with 12 stories.In this paper, the authors describe the following: the accident; the preliminary assessment of

structural conditions; the immediate intervention; the strategy adopted to consolidate the damagedstructure; the repair and strengthening works; the loading procedure of the rebuilt part of the structure;and the finishing operations. Some final remarks are also presented, including a proposal for robustnessanalysis.

© 2010 Elsevier Ltd. All rights reserved.

1. Introduction

In Portugal, the years between 1970 and 1995 were character-ized by a considerably low quality of the construction sector. Themain reasons contributing to this reality are the demographic mi-gration from rural to urban areas that started in the beginning ofthis period and the 1974 revolution that gave rise to the presentPortuguese democratic system. These two situations combined to-gether led to an abnormal increase of construction associated toan equally peculiar reduction of quality standards in constructionmaterials as well as in construction methods.At that time, in Portugal, residential buildings were generally

designed and built adopting a structure of precast prestressedconcrete beam and hollow clay block floors, with a cast-in-placeconcrete compressive layer, supported by reinforced concrete (RC)plane frames. Non-structural clay masonry walls were used bothas partitioning walls and as external perimeter walls. In the lattersituation, these were built as cavity walls, being therefore muchthicker than the inner walls.

∗ Corresponding address: ISISE – Institute for Sustainability and Innovation inStructural Engineering, Faculty of Sciences and Technology, Department of CivilEngineering, Rua Luis Reis Santos(Polo II), 3030-788 Coimbra, Portugal. Tel.: +351239797258; fax: +351 239797259.E-mail address: ejulio@dec.uc.pt (E. Júlio).

0141-0296/$ – see front matter© 2010 Elsevier Ltd. All rights reserved.doi:10.1016/j.engstruct.2010.02.018

As regards the structural design, it can be stated that Portuguesecodes have always been of high quality, at least for concretestructures. This is due to an important research activity developedat the National Laboratory of Civil Engineering (LNEC) and also tothe fact that Portuguese codes for reinforced concrete structuresare generally based on CEB provisions.This paper describes the response of a residential building

erected in the beginning of the 1980s in Coimbra, Portugal,subjected to an unforeseen event—the impact caused by alandslide, and describes how it was retrofitted. Lastly, someremarks are presented and a possible approach to enhance therobustness of this type of building is proposed.

2. Description of the accident

The year 2000 was an unusually rainy year. Just three daysbefore the new millennium, at 19:00 h, a substantial landslideoccurred, causing severe damage in the RC structure of a 16-storyresidential building, erected in the beginning of the 1980s inCoimbra, Portugal (Fig. 1(a)). The first two levels of three columnswere completely destroyed and, as a result, the rear body ofthe building supported by these, with a dimension in plant of9.5 × 6.7 m2, became a 7.0 m span cantilever with 12 stories(Fig. 1(b)). However, the damage could have been much moresevere if the flow of that significant mass of soil and debris hadnot been damped by part of the 2-story parking garage located onthe building’s backyard that completely vanished (Fig. 1(a)).

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Fig. 1. (a) Rear (West) façade of the building, a few days after the accident. (b) Detail of the total collapse of the outer columns of the damaged rear body of the building.

Fig. 2. Schematic drawing of the strut–tie system that materialized after theaccident.

The building was evacuated a few hours after the accident and,at 12:00 h the next day, a visual inspection was carried out, mainlyfocused on the damaged rear body. On the outer masonry walls,no significant anomalies were identified. Inside, a few thin cracksdistributed on the masonry walls were observed and larger cracks,with amaximumwidth of 2mm, concentrated at the corners of theopenings, were detected.

3. Preliminary assessment of the structural conditions

The observed low level of damage was attributed to the jointbehavior of the RC structure with the outer (non-structural)masonry walls that allowed a strut–tie system to materializein order to resist the gravity loads. More specifically, in thedamaged part of the building, the loads previously supportedby the destroyed columns became equilibrated by compressionstresses (struts) at the outermasonry walls and by tension stresses(ties) at the slabs (Fig. 2). In the remaining part of the building,this system originated: a tension resultant force at the top slab; acompression resultant force at the bottom slab; and an additionalcompression at the existing foundations (Fig. 2).In order to assess the safety of the damaged body, it was

necessary to quantify the stress state in the resulting structural

system. This represented a major difficulty due to the existence oftwo sets of openings in both lateral walls. Therefore, and since afast response was requested, it was decided to build a plane linearelastic finite elements model of these walls, including the columnscommon to the rest of the building. The RC structure was modeledusing linear elements and the masonry walls were simulatedwith shell elements assuming adequate geometric and materialproperties. Namely, the North outer masonry walls, presentinglarger openings, were assumed to be made of two 25 mm layersof mortar and a 70 mm layer of solid equivalent clay bricks, thelatter corresponding to the effective width of the clay bricks, i.e.,excluding the voids. To this element, with a total thickness of120 mm, an equivalent Young’s modulus was attributed, basedon an experimental study [1] conducted with similar mortar andclay bricks. Given the facts that (a) for mortar, a Young’s modulusof 3 GPa was found and, for clay bricks, the corresponding valuewas 10 GPa; (b) extrapolating these values from the mentionedresearch study [1] to the present case study comprises someuncertainty; and (c) itwould be useful to evaluate the sensitivenessof results to this parameter, it was decided to conduct a parametricanalysis, varying the equivalent Young’s modulus between 3 and10 GPa. The effect of the remaining part of the building on themodel was considered, assuming the nodes at the boundary to behorizontally restrained.From the 2D numerical analysis (Fig. 3(a)), it was possible

to conclude that (1) the finite element approach validated thehypothesis of the strut–tie model behavior; (2) the stress state inthe lateral masonry walls presented values ranging from −2 MPato 2 MPa; (3) at the lower levels, concentration zones of principaltension stresses appeared at the corners of the openings withvalues up to 4 MPa; and (4) the prediction of the maximumdeflection at the bottom levelwas between3 and8mm. Since theseresultswere in agreementwithwhatwas observed on site, this firstapproach was considered valid and the immediate interventionwas planned based on it. Later, 3D finite element models were alsodeveloped and results (Fig. 3(b)) corroborated those of this first 2Dmodeling, although with some minor differences.The original project was also analyzed. It is relevant to mention

that actions were quantified according to the 1961 RSEP [2], thePortuguese code on actions for buildings and bridges, and tothe 1967 REBA [3], the Portuguese code on reinforced concretestructures, based on the 1963 CEB ‘‘Guidelines for Contractors’’ [4].These codes already considered ultimate and service limit states asdesign criteria, instead of the traditional criteria, based on safetystresses. Nevertheless, the following relevant differences betweenthese codes and the correspondingmodern Eurocodes (EC), namelyEC 0 [5], EC 1 [6], EC 2 [7] and EC 8 [8], can be identified in thescope of the present case study: (1) the combinations of actions

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Fig. 3. Stress results of the finite element analysis. (a) 2D model: Minimum principal stresses, on the left, and maximum principal stresses, on the right, both in MPa. (b) 3Dmodel: Maximum principal stresses in MPa.

for seismic design situations according to RSEP [2] lead to lowervalues (approximately 50%) compared to EC 8 [8]; nevertheless,it should be mentioned that, in the original project, it is statedthat this situation was considered to be more unfavorable thanthe one having the wind as the leading variable action; and(2) second-order effects are considered in REBA only locally, i.e.,without considering the overall sway behavior of the structure [3];thus columns would present higher reinforcement ratios, at least,if designed according to EC 2 [7]. Moreover, concerning thebuilding’s erection, from the observation of damaged elements itwas concluded that the connections were poorly accomplishedand, although in the original project solid slabs were adopted, thecontractor decided to replace these by prestressed concrete beamand hollow clay block floors.

4. Immediate intervention

First of all, it should be stressed that the accident described hada major psychological impact on Coimbra’s inhabitants, in general,and on the building’s residents, in particular. Therefore, there wasa considerable pressure from public opinion to have a fast andreliable retrofitting intervention.The 2D numerical analysis indicated that the structure was not

at risk of imminent collapse. As mentioned above, at the outermasonry walls maximum tension stresses of approximately 4 MPawere located at the openings’ corners (where cracks appeared)andmaximum compression stresses ranging between 3 and 5MPawere identified. In the study [1] mentioned above, façade wallssimilar to the walls of the building analyzed here were tested incompression until failure, reaching ultimate values in the orderof 13 MPa. Although the acting forces in these two cases didnot have the same direction, given the significant differences inthe corresponding values, the safety of the walls was assumed.Moreover, it was also assumed that the access to the damagedzones, corresponding to approximately more 15 kN per story (5%load increase), was still admissible concerning the walls.In relation to the slabs, given the need to have a fast solution

and since the 3Dmodel was being built, the safety checkwas basedon simple calculations. According to these, the maximum tensionresultant at the slabs was approximately 600 kN. Considering that

the ‘‘ties’’ (of the strut–tie system) were materialized by 1 m ofslab width (thus by three precast prestressed concrete beams)below each outer wall combined with two peripheral RC beams,and taking into account a computed axial strength reserve of,respectively, 2×3×30 = 180 kN and 2×350 = 700 kN, totalizing880 kN, the safety of the slabs was also assumed.Finally, regarding the foundations, it was estimated that with

the accident the outer columns of the main body of the buildingbecame subjected to a 50% load increase. Nevertheless, accordingto the RC code effective when the building was designed, REBA [3],the characteristic values of both self-weight and imposed loadsare multiplied by 1.5 in a fundamental combination of actions andvariable action is considered in all stories. For this reason, followingthe accident, neither these columns nor the respective foundationswere submitted to loads above the design values. Furthermore, thevisual inspection conducted immediately after the accident did notreveal signs of settlements at the foundations or damage at thewalls.Following the analysis described above, it was decided to

undertake an immediate intervention from the inside, consisting inapplying a provisional shoring system, aiming to reduce or at leastto hold up the increase of the stress state of the masonry walls.This procedure also presented the advantage of being extremelysimple. Extensible steel post shores were applied at some pointsin the interior and wooden bars were positioned in the openings(Fig. 4(a)).Time evolution of the cracks’ widths was registered at some

locations, selected between those presenting a pattern morein agreement with the new structural system. As an example,Fig. 4(b) shows the crack at the corner of the window presented inFig. 4(a). The corresponding daily record of the crackwidthwas thefollowing: 1.0 mm, at 11:30 h on December 29; 1.0 mm, at 14:00 hon December 30; and 1.0 mm, at 16:00 h on December 31. In thiscase, as in the remaining cases, it was observed that major cracksdid not increase. In fact, only new and much smaller cracks couldbe noticed with time. For this reason, and also for security reasons,it was decided to stop monitoring cracks, just three days after theaccident.At the outer walls, all cracks were painted (Fig. 5). Mapping the

cracks had the major goal of facilitating the localization of theseafter the retrofitting operations, i.e., after closing them.

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Fig. 4. (a) Provisional shoring system applied inside. (b) Crack width monitoring during the first three days after the accident.

Fig. 5. Mapping cracks on the outer walls.

5. Strategy adopted for consolidation

The immediate intervention referred to represented onlya residual increase on the structure safety. Therefore, it wasdecided to proceed with a more substantial shoring organizedin two phases. First, a self-equilibrated prestressing system wasconceived and applied with the intention of consolidating thedamaged body of the building, thus allowing the safe removal ofdebris in the accident zone. Then,with the area below the12 storiesbeing cantilever free, a second shoring system materialized withfive steel shores was planned and applied.The prestressing system was intended to suspend the bottom

part of the cantilever from the top level of the damaged rear

body using prestressing strands (Fig. 6(a)). This solution, besidesits simplicity, presented the additional advantage of being self-equilibrated. In fact, it was relevant not to increase loads at thefoundations since reliable data concerning the building footingswas not available. A steel beamwas placed on the top slab and twoprestressing strandswere anchored to it (Fig. 6(b)) with a prestressforce of approximately 300 kN. This value is mainly justified bythe urgency in applying the designed consolidating system, sincethe two S1600/1800 prestressing strands with a cross-section of1.5 cm2 were adopted since they were available. Furthermore, themanufacturer recommended that the applied stress did not exceed1000 MPa, thus giving 300 kN. It should be added that minimizingstresses on the top story’s slab was also a concern. For this reason,extensible steel post shores were also applied on both storiesbelow, aiming to distribute the linear load between these three topstories. At both bottom corners of the cantilever, steel deviationdevices, specifically designed for this end, were placed to changethe strand direction (Fig. 6(c)). These devices were covered witha polytetrafluoroethylene (PTFE) layer to ensure that no prestresslosses occurred during the application of the prestress force andalso to avoid the introduction of lateral forces at these devices.With the prestressing system applied, the mass of soil and

debris was removed from the accident zone and it was possibleto execute the footings of the five steel shores and to put these inplace (Fig. 7(a)). Afterwards, since an active shoring was wanted,a prestress force of approximately 150 kN was applied to each ofthese using hydraulic jacks (Fig. 7(b)). In this case, this value waschosen since 5 MPa was assumed as an acceptable upper limit forthe stress increase locally introduced in the concrete, the loadingarea being 0.10 × 0.30 m2. It should be stressed that the buildingwas erected in a period characterized by a low quality level, as

Fig. 6. (a) Prestressing consolidation system. (b) Steel beam on top of the building and anchorage of prestressing tendons. (c) Steel deviation device.

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Fig. 7. (a) Provisional steel shores. (b) Prestressing the shores with hydraulic jacks.

Fig. 8. (a) Foundation for the retrofitting steel frame and shear wall. (b) Assembling the retrofitting steel frame. (c) Additional bracing steel shores.

mentioned before; the structure was damaged by the accident;and, according to the project, a B225 concrete (fck ≈ 18 MPa) wasadopted. Lastly, a triangular steel bracing system was welded tothe steel shores (Fig. 8(a)).

6. Repair and strengthening

Once the consolidation operations of the damaged body ofthe building had been concluded, it was possible to move onwith the rehabilitation work of both structural and non-structuralelements. Since these operations were intended to take placeas quickly as possible, and also aiming to minimize the time-dependent deformations of members to rebuild, a composite steeland concrete solution was adopted.Instead of the original isolated footings, it was decided to

adopt a combined footing for the three columns, aiming toobtain a uniform stress diagram at the soil/footing interface.Assuming a conservative design, a considerable contact area(11.5 × 3.0 m2) was adopted with the purpose of minimizingthe stresses and consequently minimizing the time-dependentsettlements. According to the geotechnical study referred to in theoriginal project, later corroborated by an expert who collaboratedin the retrofitting operation as a consultant, the building is foundedon a hard sandstone soil, called Grés de Silves, typical in someregions of the Iberian peninsula. This soil presents an allowablepressure of 400 kPa but, for the reasons mentioned above, a designvalue of 200 kPawas assumed. The footingwas executedwith steelbolts on top to receive the retrofitting steel frame (Fig. 8(a)).The bottom rear masonry wall of the 12-story cantilever was

replaced by an RC shear wall aiming to better accommodatethe load transfer from both provisional shoring systems to the

retrofitting steel frame (Fig. 8(a)). Inside, near the columns, steelshores were also applied to strengthen these zones.Afterward, the retrofitting steel frame was assembled and

connected to the foundation using grout to fill the voids (Fig. 8(b)).Provisional steel diagonals were welded to the retrofitting steelframe to serve as a bracing system during load transfer (Fig. 8(b)).In order to ensure an effective bracing also in the direction normalto the plane of the frame, three additional steel shores were alsolinked to the retrofitting steel frame (Fig. 8(c)).

7. Load transferring

The connection between the retrofitting steel frame and thedamaged body of the building was materialized by first applyingan interface steel beam to the bottom of the latter with epoxy-bonded steel bolts and by filing the voids also with an epoxyresin. However, this operation could not be undertaken with theprestressing system installed. Therefore, first, the prestress force ofeach of the five steel shores of the second shoring system had to beslightly increased. Then, the prestressing strands and the deviationdeviceswere removed. And, lastly, the interface steel beamwas putin place.Provisional steel corbels were welded to the retrofitting steel

frame on both sides of each column (Fig. 9). The aim was totransfer the load supported by the provisional shoring system tothe retrofitting steel frame in one single step, using simultaneouslysix hydraulic jacks placed on top of these corbels (Fig. 9). Linkingsteel elements, specially designed to be placed between theinterface steel beam (connected to the cantilever bottom) and theretrofitting steel frame, were then supposed to be welded to both

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Fig. 9. Schematic drawing of load transfer using hydraulic jacks on top of provisional steel corbels, welded to each of the retrofitting columns, and linking steel element tobe placed between and welded to both the retrofitting column and the interface steel beam.

Fig. 10. Set up for load transfer using hydraulic jacks.

(Fig. 9). Nevertheless, since the contractor was only able to providetwo hydraulic jacks, the load transfer was performed in severalsteps instead, using steel pads to be able to move the hydraulicjacks fromone corbel to another andwelding the three linking steelelements finally (Fig. 10).

8. Finishing operations

After mobilizing the retrofitting steel frame, all steel shoresfrom both the bracing system and the provisional shoring systemwere removed, as well as the steel corbels. Since, in the case of anearthquake, tension stresses between the retrofitting steel frameand the original structure will appear, it was decided to strengthenthis connection. With this aim, transfer steel plates were weldedto the interface steel beam and connected to each of the three RCcolumns with epoxy-bonded steel bolts (Fig. 11).The destroyed slab was rebuilt adopting a solid concrete slab

supported by steel beams normal to the plane of the steel frame(Fig. 12(a)). In this process, steel elements from the provisionalsystems were reused. Next, all steel members were provided withsteel reinforcement and covered with a high-performance grout(Fig. 12(b)) aiming to ensure effective fire and corrosion protectionand also assumed as an additional strengthening measure.Finally, non-structural finishing operations were carried out

and the building was painted (Fig. 12(b)). Later, the 2-storyparking garage located in the backyard that had vanished was

a b

Fig. 11. (a) Schematic drawing of the transfer steel plates welded to the interfacesteel beam and connected with M20 steel anchors to each of the existing RCcolumns. (b) Detail of the transfer steel plates.

Fig. 12. (a) Retrofitting steel frame after removing the provisional systems andrebuilding the destroyed slab. (b) Retrofitting operations concluded.

also rebuilt (Fig. 13) and the building retrofitting was completelyaccomplished.

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Fig. 13. Rebuilt parking garage.

9. Final remarks

Relative to the intervention, it should be noted that thedesign of all structural members was performed consideringthe construction phases described. Consequently, the retrofittedstructure turned out to be considerably more resistant than theoriginal one. As an example, it can be mentioned that the loadtransfer phase constituted amore unfavorable loading situation, interms of stresses and supporting conditions, than the final serviceconditions. Furthermore, it was ensured that the connectionbetween the new and the existing structural members wascompatible with the maximum strength capacity of these.As regards the original structure, it can be stated that it

corresponds to a typical residential building erected in Portugalin the time period referred to in the introduction. In fact, thematerials and methods at that time were not outstanding, for thereasons given, but, nevertheless, the building had a satisfactoryresponse to the unforeseen event of an impact of a mass of soiland debris against its RC structure. For this reason, it can be statedthat the structure of this building is robust, since it was capableof preventing a progressive collapse after a localized damage. Thisproperty, robustness, gained much interest after the Ronan Pointblock of flats disaster in 1968, caused by an explosion, and morerecently after the Twin Towers catastrophe in 2001, caused by aterrorist attack.Different researchers have proposed different approaches to

achieve structural robustness, including the creation of alternate

load paths in the structure; the improvement of redundancy andductility; and the reduction of risk of abnormal loads, throughprotection of the structure [9]. Nevertheless, this concept is notyet included in codes, at least not in a helpful way. Taking theEurocodes as an example, only in EC 1 [6] (part 7) does a definitionof robustness appear: ‘‘the ability of a structure to withstandevents like fire, explosions, impact or the consequences of humanerror, without being damaged to an extent disproportionate tothe original cause’’, although no practical measures are provided.Some organizations have recently published some guidelines onrobustness, such as IABSE [10], and there is a COST Action inprogress to study this issue [11]. Nonetheless, common to allproposed approaches is the fact that these are always focusedonly on the structure. Following the experience described here, theauthors would like to present a different suggestion concerningthis subject.The building described in this paper was able to withstand the

mentioned unforeseen event only because both lateral masonrywalls were mobilized. It seems therefore rather interestingto start considering non-structural elements as a structuralreserve, for this scenario only, not for fundamental, accidental orseismic combinations of actions. Naturally, specifications on thematerial properties and on the construction procedures, includingdetailing guidelines, of these non-structural elements will also berequired.

References

[1] Vicente R. Pathology of facadewalls—Mechanical behavior of facadewallswithexternal correction of thermal bridges. M.Sc. thesis. University of Coimbra.2002 [in Portuguese].

[2] RSEP. Code on actions for buildings and bridges. 1961 [in Portuguese].[3] REBA. Code on reinforced concrete structures. 1967 [in Portuguese].[4] Recommendations pratiques à l’usage des constructeurs. Comité Européen duBéton. 1963.

[5] Eurocode 0: Basis of structural design. European Committee for Standardiza-tion. 2002.

[6] Eurocode 1: Actions on structures. European Committee for Standardization.2006.

[7] Eurocode 2: Design of concrete structures. European Committee for Standard-ization. 2004.

[8] Eurocode 8: Design of structures for earthquake resistance. EuropeanCommittee for Standardization. 2004.

[9] Ellingwood BR, Dusenberry DO. Building design for abnormal loads andprogressive collapse. Comput Aid Civ Inf Eng 2005;20:194–205.

[10] Knoll F, Vogel T. Design for robustness, structural engineering documents.Internat Assoc Bridge Struct Eng 2009.

[11] COST-TU0601. Memorandum of understanding. 2007.