Soil Dynamics and Earthquake Engineering 42 (2012) 229–245

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The Maule (Chile) earthquake of February 27, 2010: Development of hazard,site specific ground motions and back-analysis of structures

Amr S. Elnashai a, Bora Gencturk n,b, Oh-Sung Kwon c, Youssef M.A. Hashash a, Sung Jig Kim d,Seong-Hoon Jeong e, Jazalyn Dukes f

a Department of Civil and Environmental Engineering, University of Illinois at Urbana-Champaign, Newmark Civil Engineering Lab, 205 N Mathews Ave, Urbana, IL 61801, USAb Department of Civil and Environmental Engineering, University of Houston, N107 Engineering Building 1, Houston, TX 77204-4003, USAc Department of Civil Engineering, University of Toronto, 35 St. George St., Toronto, ON, Canada, M5S 1A4d Department of Architectural Engineering, Keimyung University, 1095 Dalgubeoldaero, Dalseo-Gu, Daegu 704-701, South Koreae Department of Architectural Engineering, #2S413, Inha University, 253 Yonghyun-Dong, Nam-Gu Incheon, 402-751, South Koreaf School of Civil and Environmental Engineering, Georgia Institute of Technology, 790 Atlantic Drive, Atlanta, GA 30332, USA

a r t i c l e i n f o

Article history:

Received 16 February 2011

Received in revised form

21 February 2012

Accepted 8 June 2012Available online 13 July 2012

61/$ - see front matter & 2012 Elsevier Ltd. A

x.doi.org/10.1016/j.soildyn.2012.06.010

esponding author. Tel.: þ1 713 743 4091; fax

ail address: bgencturk@uh.edu (B. Gencturk).

a b s r a t c t

The Maule (Chile) earthquake of 27 February, 2010 has caused severe disruption and economic losses.

With a magnitude of 8.8, it has been recorded as one of the largest earthquakes of the last century. The

ground motion records from large subduction earthquakes, such as the Chile earthquake, are sparse.

The number of accelerograms that recorded the strong ground motion was relatively few and only a

few of these ground motions were released to engineering community. One of the objectives of this

paper is to develop site specific ground motions that take into account the particular characteristics of

this major earthquake. These are proposed to the engineering community as representative ground

motions based on the best available data. The second objective of the paper is to investigate, using

numerical tools, some typical failures observed in the engineered buildings and bridges. Although, in

general engineered structures performed very well and the majority of failures, hence losses, were to

non-engineered structures, some repeated deficiencies in structural design were observed. The

developed hazard and site specific ground motions are used as inputs for inelastic dynamic analysis

of advanced finite element building and bridge models. The results are processed to explain

quantitatively the structural deficiencies observed in the field.

& 2012 Elsevier Ltd. All rights reserved.

1. Preamble

On February 27, 2010 at 03:34 am local time, a powerfulearthquake of magnitude 8.8 struck central Chile. The epicenter ofthe earthquake was approximately 8 km off the central region ofthe Chilean coast. With an inclined rupture area of more than80,000 square km that extends onshore, the region of Maule wassubjected to a direct hit, with an intense shaking duration of atleast 100 s, and peak horizontal and vertical ground acceleration ofover 0.6 g. Over 800,000 individuals were directly affected throughdeath, injury and displacement. According to the Ministry ofInterior of Chile, the earthquake caused the death of 521 persons,with almost half of the fatalities caused by the consequentialtsunami. More than a third of a million buildings were damaged tovarying degrees, including several cases of total collapse of majorstructures. The transportation system was dealt a crippling blow,

ll rights reserved.

: þ1 713 743 4260.

with 830 failures registered with the Ministry of Public Works onroads in both the public and private transportation networks. Outof 130 hospitals in the effected region, four became uninhabitable,12 had greater than 75 percent loss of function, only eight werepartially operational after the main shock, and 80 hospitals neededrepairs. A total of 4013 schools (representing nearly half of theschools in the affected areas) suffered significant damage. Severedisruption of commerce as well as the rescue and response effortresulted from the damage to roads, embankments, bridges, portsand airports. According to the Ministry of Treasury, the economiclosses are estimated to be $30 billion (loss of infrastructure: $20.9billion, loss of production: $7.6 billion, other costs such asnutrition and debris removal: $1.1 billion) which is equivalent toapproximately 17 percent of the GDP of Chile.

Only a few acceleration records were released to the engineeringcommunity as of January 2011; this withholding of such infor-mation of great importance to detailed studies that benefitssociety at-large is regrettable. The available records confirm thesevere shaking that resulted from the earthquake and the longduration of the strong-motion part of the records. The dearth of

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245230

available records led the authors to generate spectrum compatiblesignals. Candidate attenuation relationships that are suitable forlarge subduction earthquakes are selected and amongst these, theones that give the best correlation to the available data areidentified. The identified attenuation relationships are combined todevelop site specific hazard spectra. In order to generate accelera-tion time histories for back-analysis of a building and two bridges,ground motions are selected from the available digital recordingsand spectrum matching is utilized to obtain the site specific timehistories at the locations of the back-analysis structures.

The objective of conducting back-analysis of structures is toexplain some typical failures to engineered structures observed inthe field. The selected building is a three story flat-slab reinforcedconcrete structure and the back-analysis aims to explain theparticular damage pattern observed only at the first story col-umns of the exterior frames of the buildings. The most commonlyobserved bridge damage during the Chile earthquake was due tounseating or displacement of superstructure, especially forskewed bridges. This resulted in severe yielding of seismicrestrainers, which are extensively used in bridge construction inChile as opposed to other parts of the world, and the destructionof shear keys in abutments. One of the bridge case studies focuseson explaining the effect of seismic restrainers on the bridgeperformance. And the other bridge case study investigates thereasons for excessive displacements of the superstructure thatresulted in unseating and collapse of several bridges. A verytypical bridge has been selected for the second case study andseismic fragility functions are also developed using hazard spe-cific ground motions from different parts of Chile.

Fig. 1. Recent major earthquakes in Chile and seismic gaps, reproduced from [6].

2. Engineering seismology

2.1. Seismo-tectonic environment

The Maule earthquake struck Chile on 27 February 2010 at 03:34a.m. local time. The magnitude of the earthquake is estimated as Mw8.8. The epicenter was located offshore at 35.9091S, 72.7331W withthe following distances to major cities: Chillan 95 km, Concepcion105 km, Talca 115 km, and Santiago 335 km. The hypocenter was35 km deep [27]. The average slip over the approximately 81,500 km2

rupture area was 5 m, with slip concentrations down-dip, up-dip andsouthwest, and up-dip and north of the hypocenter. Relatively littleslip was observed up-dip/offshore of the hypocenter. The averagerupture velocity was estimated to be in the range of 2.0–2.5 km/s. TheGlobal Centroid Moment Tensor (GCMT) solution yielded a seismicmoment of 1.84�1022 Nm, a centroid location of 35.951S, 73.151W,and a best double couple fault plane geometry with strike and dipangles of 181, and a rake angle of 1121 [14]. The size of the fault zonevaries depending on the calculation method. According to methodproposed by the National Earthquake Information Center (NEIC), thesize of the fault zone is 189�530 km. The slip amplitude reached9 m at peak location. As a comparison, the maximum value of slip inthe 2004 Sumatra and recent Haiti earthquakes are about 20 m and5 m respectively. The uplift reached as high as 2 m and settlements of0.4 m were observed. The coast of Chile moved west, into the oceanas much as 6 m at some locations.

The earthquake nucleated on the subduction zone that runsalong the entire �5000 km length of the western coastline ofSouth America, known as the Peru–Chile trench. Earthquakes inthis region are due to stress buildup resulting from the movementof the oceanic Nazca plate eastward and downward towardsthe South American plate at a rate of approximately 70 mm peryear [25].

Due to the close proximity to the Nazca-South Americasubduction zone, Chile has long been subjected to earthquakes

of large magnitude. On average, a magnitude 8.0 earthquakeoccurs every decade and a magnitude 8.7 earthquake or greateris observed within a century. The map in Fig. 1 shows recentmajor earthquakes in Chile alongside seismic gaps. The whitecircles indicate the rupture for individual events, the red circlesshow the epicenters and the yellow dots are the aftershocks. Fromnorth to south, the Peru–Chile trench has ruptured in severalincidents except for three locations that are shown with red lines.The segment of the subduction zone, Concepcion gap [24] inFig. 1, between the 1985 Valparaıso and the 1960 Valdivia earth-quakes, which last produced an earthquake in 1835, is the zonethat ruptured during the 2010 earthquake. Another segment, theArica gap has been relatively inactive since 1877 and has thepotential to produce an earthquake of magnitude 8.0 to 8.5. Therehas been energy release with the Valparaıso (1906 and 1985), LaSerena (1943) and Vallenar (1922) earthquakes in the La Serenaseismic gap but it is uncertain if all of the energy accumulated inthis region has been released with these events.

2.2. Shaking intensity and recorded strong ground motion

Ground motions were recorded by two departments at theUniversity of Chile. At the time of writing this paper, digitalrecordings from 10 stations were available through the

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245 231

Seismological Service at the Department of Geophysics. Addition-ally, plots of recordings (both acceleration time histories andresponse spectra) from 9 stations were available through RedNacional de Acelerografos Departamento de Ingenieria Civil(RENADIC). No information was available regarding the soilproperties where the stations are located. The recordings arelisted in Table 1, along with peak ground acceleration of eachcomponent, distances to fault based on different measures andmaximum duration of horizontal components. The finite faultmodel by U.S. Geological Survey [26] is used to calculate theclosest distance to the rupture plane. As described earlier, the sizeof the fault zone is 189�530 km and the fault strike, dip and rakeangles are approximately 17.51, 181 and 1121, respectively. Thedurations are based on the bracketed (with a threshold¼0.05 g)and significant duration definitions. Fig 2(a) shows the surfaceprojection of the fault and the location of the stations alongwith zones defined by the Chilean seismic code, NCh 433 [8].Fig 2(b) plots the acceleration time histories and the respectiveArias intensities for the station labeled as CCSP.

As given in Table 1, the peak ground accelerations (PGA) of thehorizontal components from CCSP and MELP stations are 0.65 gand 0.78 g, respectively. These are the highest PGAs recordedduring the earthquake. The significant duration of CCSP record,based on Arias intensity, is 67.6 s. The bracketed duration, with athreshold of 0.05 g, is much longer, yielding 113 s of strongground shaking for the CCSP record. The existing attenuationrelationships for strong motion duration [10,22] show that thebracketed duration decreases with increasing distance from thesource, while the significant duration increases. This is becausethe bracketed duration uses the absolute threshold of the ampli-tude, while the significant duration utilizes the relative thresholdand thus, is related to the geometry of the accelerogram, regard-less of its absolute amplitude. Therefore, to compare the durationof the strong ground motion from the Chile earthquake, thebracketed durations (with a 0.05 g threshold value) of largeearthquake records from the Pacific Earthquake EngineeringResearch (PEER), Next Generation Attenuation (NGA) projectdatabase (http://peer.berkeley.edu/nga) are utilized. Relativelylarge earthquakes (MwZ6.9) with a peak ground acceleration of

Table 1Information on stations and the recorded strong ground motions.

Station PGA (g)

ID Name NS EW Vert

Seismological serviceCCSP Colegio San Pedro, Concepcion 0.65 0.61 0.58

CSCH Casablanca 0.29 0.33 0.23

MELP Melipilla 0.57 0.78 0.39

ANTU Campus Antumapu, Santiago 0.23 0.27 0.17

STL Cerro Santa Lucia 0.24 0.34 0.24

LACH Colegio Las Americas 0.31 0.23 0.16

CLCH Cerro Calan, Santiago 0.21 0.23 0.11

OLMU Olmue 0.35 0.25 0.15

SJCH San Jose de Maipo 0.47 0.48 0.24

ROC1 Cerro El Roble 0.19 0.13 0.11

RENADICMMVM Vina del Mar (Marga Marga) 0.35 0.34 0.26

CEVM Vina del Mar (Centro) 0.22 0.33 0.19

MAIP CRS MAIPU RM 0.56 0.48 0.24

CURI Hosp. Curico 0.47 0.41 0.20

SRSA Hosp. Sotero de Rıo 0.27 0.26 0.13

UCSA Universidad de Chile, Santiago 0.17 0.16 0.14

MMSA Estacion Metro Mirador Santiago 0.24 0.17 0.13

LTSA Hosp. Luis Tisne RM 0.30 0.29 0.28

VALD Hosp. Valdivia 0.09 0.14 0.05

a Distance—de: epicentral distance, dh: Hypocentral distance, dsp: distance to surfab Duration—Db: bracketed duration with a threshold of 0.05 g, Ds: significant dura

0.1 g or more and source distances of less than 100 km areselected. The durations are compared in Fig. 3, which indicatesthat the strong ground shaking during the Chile earthquake wassignificantly longer compared to other earthquakes. The longduration of the Chile earthquake is an important feature thatneeds to be taken into account in the development of site specificground motions that are discussed in Section 2.3.

The 5 percent damped elastic spectra for the ground motions areshown in Fig. 4. The spectra are separated based on the zonesspecified in the Chilean seismic code. The recommended designspectra for soil types I to III are also shown in the plots. Note thatsoil type IV is omitted because none of the records considered hereare on such soil. Although soil type information is not available forthe recording stations, it is seen that the recorded spectra exceedthe design spectra for soil types I and II for the intermediate periodrange. Additionally, the spectra at stations MAIP, CURI, CCSP andMELP exceed significantly the design spectra for periods shorterthan 1 s. For these records, the peak spectral ordinates normalizedby individual PGAs are 4.26, 4.06, 3.39, and 3.67, respectively,which indicate the severity of the ground shaking, particularlywhen compared with the amplification factors from the Chileanseismic code, which vary from 2.76 to 3.09 depending on the soiltype. Such a feature could result in relatively high demand imposedon short period structures that are designed to conform to coderequirements. Overall, the comparison between the unnormalizedspectra of the individual records from the 2010 earthquake and thecode spectra suggests a need to amplify the latter.

The available data from the 19 stations in Chile are used to selectappropriate attenuation relationships for horizontal ground motions.Attenuation relationships that are developed for subduction zoneswith thrust mechanisms and that utilize a large and uniformlyprocessed database of large magnitude events are selected here. Fivecandidate attenuation relationships are identified: Zhao et al. [31],Atkinson and Boore [2], Campbell and Bozorgnia [3], Gregor et al. [7],and Youngs et al. [30]. Note that most other attenuation relationshipsavailable in literature are not developed to account for large magni-tude earthquakes such as the 2010 Chile earthquake. Due to the lackof soil type information for the ground motion records, rock siteattenuation relationships are used, to generate acceleration response

Distance (km)a Duration (sec)b

de dh dsp drup Db Ds

109.1 114.6 0.0 36.4 113.1 67.6

311.7 313.6 20.9 48.5 50.3 32.2

283.0 285.1 0.0 52.5 60.2 31.9

323.0 324.9 25.3 66.1 45.9 37.7

334.2 336.0 32.5 69.2 58.9 41.2

339.1 340.9 39.8 72.9 54.9 37.5

343.8 345.6 43.1 74.8 49.3 42.4

353.7 355.4 62.2 78.6 48.9 32.1

332.5 334.4 49.8 78.8 71.2 38.6

361.6 363.3 67.9 85.4 34.9 35.5

336.7 338.5 47.0 60.8 unavailable

337.8 339.6 48.5 61.0

321.3 323.2 19.1 64.0

170.4 174.0 13.0 65.1

325.2 327.1 29.7 68.0

331.6 333.5 30.0 68.0

329.5 331.3 30.3 68.2

332.3 334.2 33.4 69.6

437.8 439.2 182.8 192.7

ce projection of the fault, and drup: distance to rupture plane.

tion (5–95% of Arias Intensity).

0 10 20 30 40 50 60 70 80 90 1000

20

40

60

80

100

120

Closest Distance to Rupture Plane (km)

Bra

cket

ed D

urat

ion

(>0.

05g,

sec

) Cape Mendocino (1992), Mw= 7.0Chi-Chi, Taiwan (1999), Mw= 7.6Denali, Alaska (2002), Mw= 7.9Duzce, Turkey (1999), Mw= 7.1Hector Mine (1999), Mw= 7.1Imperial Valley (1940), Mw= 7.0Irpinia, Italy (1980), Mw= 6.9Kern County (1952), Mw= 7.4Kobe, Japan (1995), Mw= 6.9Kocaeli, Turkey (1999), Mw= 7.5Landers (1992), Mw= 7.3Loma Prieta (1989), Mw= 6.9Manjil, Iran (1990), Mw= 7.4St Elias, Alaska (1979), Mw= 7.5Tabas, Iran (1978), Mw= 7.3Chile (2010), Mw = 8.8

Fig. 3. Comparison of strong ground shaking during the Chile earthquake with other notable earthquakes.

-0.8

-0.4

0

0.4

0.8

Acc

eler

atio

n (g

) PGA: 0.65 g

-0.8

-0.4

0

0.4

0.8

Acc

eler

atio

n (g

)

PGA: 0.61 g

-0.8

-0.4

0

0.4

0.8

Acc

eler

atio

n (g

) PGA: 0.58 g

0 20 40 60 80 1000

50

100

Time (sec)

Aria

s In

tens

ity (%

)

NSEWVert

NS

EW

Vert

Fig. 2. (a) Fault plane and the seismic zones from the Chilean seismic code and the locations of the stations (yellow and red filled circles indicate records from

Seismological Service and RENADIC, respectively); (b) acceleration time histories and Arias intensity at station CCSP. (For interpretation of the references to color in this

figure legend, the reader is referred to the web version of this article.)

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245232

spectra. Relationships for rock sites are observed to provide the bestfit to the available data points (see Fig. 5). The predictions of thecandidate attenuation relationships for the various soil classificationsare provided in Elnashai et al. [4]. The PGA and spectral accelerationvalues predicted by the attenuation relationship by Zhao et al. [31]and Campbell and Bozorgnia [3] correlate best with the measureddata (see Fig. 5). Therefore, these attenuation relationships areselected to generate the ground motions required for back-analysisof structures.

2.3. Development of suites of ground motion for structural analysis

No recordings exist at the locations of the case study structures,therefore, a suite of spectrum compatible records that are

representative of the hazard at the locations of the case studystructures are generated. Most of the stations provided by RENA-DIC are located north of the rupture plane and none of the stationsare within 100 km from the case study structures. Additionally, nodigital records from RENADIC are available. Thus, the digitalrecordings available through the Seismological Service at theUniversity of Chile are used as seed signals. Locations of the casestudy structures and stations are shown in Fig 2(a). The accelera-tion response spectra which take into account the distance fromthe fault are considered for generating the spectrum compatiblerecords. Particularly, the selected attenuation relationships arecombined accounting for their standard deviation as follows:

RSA¼ RSAzðmz7aszÞW1þRSAcðmc 7ascÞW2 ð1Þ

0.1

0.5

1

2

PG

A (g

)

0.1

0.5

1

2S

a (g

) at T

=0.2

0 se

c

10 50 100 2000.1

0.5

1

2

Sa

(g) a

t T=0

.40

sec

Closest Distance to Rupture Plane (km)10 50 100 200

0.1

0.5

1

2

Closest Distance to Rupture Plane (km)

Sa

(g) a

t T=1

.00

sec

Zhao et al. (2006) Atkinson and Boore (2003) Campbell and Bozorgnia (2003) Gregor et al. (2002)

Youngs et al. (1997) Seis Service RENADIC

Fig. 5. PGA and response spectra attenuation relationships for rock.

0 0.5 1 1.5 2 2.5 30

0.5

1

1.5

2

2.5

Period (sec)

Sa

(g)

CCSPCSCHMELPOLMUSJCHROC1MMVMCEVMVALD

0 0.5 1 1.5 2 2.5 30

0.5

1

1.5

2

2.5

Period (sec)

Sa

(g)

ANTUSTLLACHCLCHMAIPCURISRSAUCSAMMSALTSA

Zone 3Zone 2

Soil ISoil IISoil III

Soil ISoil IISoil III

Fig. 4. Comparison of measured spectra (5 percent damping) with design spectra from the Chilean seismic code; the thick and thin lines show the NS and EW components

respectively.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245 233

where: RSAz is the attenuation relationship by Zhao et al. [31], RSAc

is the attenuation relationship by Campbell and Bozorgnia et al.[3], m is the mean value, a is the weighting constant for standarddeviation, here selected as 0, 70:5, 71:0 and7 1:5, s is thestandard deviation, and W is the weight on the spectral accelera-tion attenuation relationship, here selected as 0.5.

Fig. 6 shows the comparison between the combined attenua-tion relationship according to Eq. (1) and PGA or spectral values ofthe recorded ground motions. The attenuation relationship isplotted along with selected standard deviation values of 70:5s,71:0s and 71:5s. The combined attenuation relationship isused to generate the target spectra at the case study sites. Themean spectra and spectra generated with 70:5s are selected tobe used in spectrum matching of the case study records as thisrange is considered to capture the variation in the strong groundmotions from the Chile earthquake.

The spectrum-matched records are generated by using Wav-Gen [20] program. WavGen modifies a given seed record to renderit compatible with a given spectrum. As mentioned earlier, theseed records are selected from the ground motions recorded bythe Seismology Service at the University of Chile. Table 2 lists theseed records for each site and Fig. 7 shows example spectrumcompatible acceleration time histories and alongside the targetand matched response spectra.

3. Observed structural damage

This paper focuses on the damage to a selected building andtwo bridges. Therefore, the observed damage to these structuraltypes during the 2010 Chile earthquake is very briefly summarizedin this section.

0.1

0.5

1

2P

GA

(g)

mean ±1.5σ mean ±σ mean ±0.5σ mean Seis. Service RENADIC

0.1

0.5

1

2

Sa

(g) a

t T=0

.20

sec

10 50 100 2000.1

0.5

1

2

Sa

(g) a

t T=0

.40

sec

Closest Distance to Rupture Plane (km)10 50 100 200

0.1

0.5

1

2

Closest Distance to Rupture Plane (km)S

a (g

) at T

=1.0

0 se

c

Fig. 6. Combined PGA attenuation relationship for rock site and spectral acceleration at different periods.

Table 2Seed records used for generating synthetic ground motions for back-analysis of

the case study structures.

Site Fault distance (km) Seed records

Odontology building 37.9 CCSP-NS, MELP-NS

ITATA Bridge 53.4 MELP-NS, STL-EW

Bridge on route (Ruta) 5 68.9 LACH-NS, STL-EW

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245234

3.1. Observed building damage

According to the Ministry of Housing and Urban Developmentof Chile, a total of 370,051 houses were damaged from earthquakeand the consequential tsunami. The damage distribution is givenin Table 3. The cost of damage to private houses is estimated as$3.7 billion. The adobe construction in Maule region suffered themost damage. In Curico, 90 percent of adobe construction wasdestroyed. Failure to engineered buildings was due to the com-mon causes of irregularity and limited ductility, with a few casesof damage due to a special provision in the Chilean code thatallows structural walls with thin webs. On the whole, theperformance of engineered structures was reasonable, taking intoaccount the magnitude and proximity of the earthquake. A studyby Rene Lagos using the building permit statistics from NationalInstitute of Statistics Chile indicates that out of 9974 buildingsconstructed between 1985 to 2009, only four buildings collapsedand 50 buildings need to be demolished [15]. Less than 2.5 percentof engineered structures in Chile suffered damage and out of allcasualties, less than 20 died in engineered buildings.

3.2. Observed bridge damage

Considering the large earthquake magnitude and rupture area,the damage to bridge structures was less than that could havebeen expected. The major highways in Chile are constructed andmaintained by private companies. Based on the informationobtained from researchers in Chile during the field investigations,the highway network is approximately 2200 km in length and hasaround 2000 bridges with span lengths longer than 10 m. Among

these structures, only one percent (8 highway and 12 pedestrianbridges) collapsed due to the earthquake. Approximately 100bridges (50 highway and 50 pedestrian bridges) were damagedto a level requiring repair. According to Yen et al. [29], whenshorter-span bridges (excluding culverts and pedestrian bridges)are counted, the number of collapsed or damages bridges isapproximately 3% of the total number of bridges in Chile. Duringthe field investigations the authors visited several damagedbridges to study failure modes. Additionally, there had beenseveral teams from other organizations, such as EarthquakeEngineering Research Institute (EERI) and Japan Society of CivilEngineers (JSCE), which focused specifically on bridge perfor-mance. Table 4 presents a compiled list of damaged bridges, theirlocations, and failure modes based on the data collected from thefield investigation and reports available in the public domain[11,28].

The most commonly observed bridge damage was due tounseating or displacement of superstructure, especially forskewed bridges. Even if the centers of mass and stiffness coincide,skewed bridges tend to develop rotation, which results in unseat-ing of the bridge girder or failure of shear keys. Bridges designedaccording to relatively recent design practice apparently sufferedmore damage than bridges constructed in the past. In the oldconstruction practice, the integrity of bridge superstructure washigh due to diaphragms connecting the girders. On the otherhand, bridges constructed in recent years often did not havediaphragms. As a result of the lack of in-plane stiffness andconnectivity, bridge girders were damaged due to the poundingof the superstructure onto the shear keys. In the report fromanother field investigation team [28], it is stated that the seat-width on bridge bent was not large enough to prevent unseating.In addition, in many bridges, shear keys were not strong enoughto resist the lateral forces from the pounding superstructure. Thespecifications in the United States require that the shear keys,restrainers, and bearing seat width should be properly designedsuch that the loss of bridge spans due to unseating can beprevented. Elastomeric bearings that are often used in overcross-ing bridges in Chile can only be used in single span bridges in theUnited States. For multi-span continuous bridges, bridge decks

Table 3Housing damage distribution.

Destroyed Majordamage

Minordamage

Total

Coastal 7931 8607 15,384 31,922

Urban adobe 26,038 28,153 14,869 69,060

Rural adobe 24,538 19,783 22,052 66,373

Government housingdevelopments

5489 15,015 50,955 71,459

Private housingdevelopments

17,448 37,356 76,433 131,237

TOTAL 81,444 108,914 179,693 370,051

-0.6

-0.3

0

0.3

0.6

Acc

eler

atio

n (g

)

0

0.5

1

1.5

Sa

(g)

-0.6

-0.3

0

0.3

0.6

Acc

eler

atio

n (g

)

0

0.5

1

1.5

Sa

(g)

0 10 20 30 40 50 60 70 80 90-0.6

-0.3

0

0.3

0.6

Time (sec)

Acc

eler

atio

n (g

)

0 1 2 30

0.5

1

1.5

Period (sec)

Sa

(g)

Fig. 7. Spectrum matched records for back-analysis of the case studies.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245 235

are connected to bent-caps, which prevents rotation and/orunseating of bridge decks during an earthquake. If more stringentrequirements for shear keys and bearing seat width were stipu-lated in the bridge design specifications in Chile, the damage fromunseating of superstructure could have been minimized withoutsignificantly increasing the bridge construction cost.

Only a few bridges reportedly had damage to their substructure.It is speculated that as there are not many bridges with integralsuperstructure–substructure connection, the large inertial forcefrom superstructure was not transferred to the substructure.The inertial force caused the unseating or large movement of thesuperstructure, or the failure of shear keys.

4. Case studies

In this section, a building and two bridges are modeled andanalyzed using finite element software under the site-specificground motions described in Section 2.3. The numerical modelsare based on complete design drawings and measurementsobtained during the field investigations. The objective here is tobuild on the field observations and identify the sources (deficien-cies in structural designs) that contributed to the observed

failures. The damage to the building was peculiar in that it waslimited to the first story columns of the exterior frames. Thedamage mainly resulted from the short column effect due to theexistence of masonry infill walls which most probably were nottaken into account in the design. This is a typical failure modeobserved in most earthquakes. One bridge case study aims atexplaining the effect of seismic restrainers, which are commonlyused in Chile, on the dynamic response and the bridge selected forthe other case study exemplifies the most commonly used high-way overpass design in Chile in the recent years.Several bridges of similar type suffered damage to differentextents in the 2010 Maule earthquake. Seismic fragility functionsare also developed for this typical bridge to be used as a means topredict damage for similar structures under various earthquakeintensities.

4.1. Odontology building of the University of Concepcion

4.1.1. Introduction and building configuration

The Odontology building of the University of Concepcion isinvestigated in this case study. The building is located at thesouthern east part of Concepcion as shown in Fig. 8(a).The structure was used as the school of dentistry which alsoincluded rooms for examination and medical treatment ofpatients. Due to the requirements of the building occupancy,the reinforced concrete structure has a particular configuration.As shown in Fig. 8(b), the building has three regular stories andthree, approximately half height, service stories that serve toutility lines.

A typical floor plan of the structure is shown in Fig. 9. Thebuilding has a rectangular shape with seven bays in x-directionand five bays in y-direction. Structural core-walls are provided atthe center of the building to resist the majority of the lateralloads. Except for minor differences in column dimensions of theexterior frames that have five bays, the structure is symmetricalwith respect to both x and y axes.

Fig. 8. (a) Location of the Odontology building of the University of Concepcion and the ground motion recording station CCSP, (b) the Odontology building of the University

of Concepcion.

Table 4List of damaged and/or collapsed bridges.

Bridge name9Roadway carried by the bridge9Location Damage/failure modes and other remarks

Miraflores Bridge9Vespucio NorteExpress9Independencia, Santiago

North bridge had diaphragm between girders and concrete shear keys preventing the movement of the

superstructure. Damage on concrete shear keys. South bridge had steel shear keys for each girder. Most shear

keys were damaged.

Lo Echevers Bridge9Vespucio Norte Express9LoEchevers, Santiago

Skewed bridge. Unseating of superstructure due to large transverse displacement.

Quilicura Bridge9n/a9n/a Skewed steel girder bridge. Unseating of deck. Failure of shear keys. Shear keys at only one side was

damaged.

Las Mercedes Bridge9Route 5 South93414.32350S70145.7530W

Skewed pre-stressed girder bridge. Unseating of deck. No diaphragms joining pre-stressed girders.

Costanera Norte9n/a9n/a Skewed bridge. Shear key failure. Large displacement of superstructure at abutment.

Pasarela Peatonal9Route 5, Norte9n/a Pedestrian bridge. Bridge deck was bolted to the bent of the bridge. The bolts were sheared and the

superstructure unseated. The bridge bent was undamaged.

Perquilauquen Bridge9Route 5936115.2080S71148.82570W

Consists of two bridges, one built in mid-1990s and the other one recently built. The older bridges suffered

less damage than the new one. The damage on the new bridge was large transverse displacement of

superstructure. The shear key of the new bridge was very weak and failed.

Juan Pablo II Bridge9n/a9Concepcion, 36149.4010S7315.4770W

Shear failure on bridge column due to lateral spreading of soil toward river.

Llacolen Bridge9n/a9Concepcion, 36150.0390S7314.6230W

Indication of lateral spreading toward river. Unseating of several ramps.

Bıobıo Bridge9n/a9Concepcion, 361 50.4420S 7314.1150W

Built in 1930s. Closed before the earthquake for maintenance. Steel stringer bridge. Unseating of bridge

decks. Total collapse.

Las Ballenas Bridge9n/a9Suburb of Concepcion Rupture of elastomeric bearings.

Rıo Claro Bridge9n/a9n/a Unreinforced masonry bridge built in 1870. Collapsed during the earthquake.

Tubul Bridge9n/a9Arauco South most location of complete bridge collapse. Steel girder bridge. All eight steel girders were unseated. In

Ref. [11], it was reported that performance of foundation was insufficient.

Paso Cladio Arrau9n/a9361 39.5360 S 721 19.5450W Minor transverse translation of the superstructure. The bridge was serviceable after the earthquake.

Route 5 overpass near Chillan9Route 5(overpass)9Chillan 361 35.0830S 721 6.5340W

Skewed bridge. Large translation and close to unseating of superstructure. Shear key at the abutment was not

long enough to provide resistance to the superstructure.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245236

As shown in Fig. 10, three floors of the building are designed asflat slabs. Another feature of the building is that masonry infill wallsare used at the first floors of the exterior frames. Based on theavailable information, it is not certain whether the lateral resistanceprovided by these infill walls is included in design calculations. It ispostulated that these walls were considered to be non-structural.The height of the first story is 3.1 m and the height of the masonryinfill walls was measured to be approximately 2.1 m.

4.1.2. Observed damage

The earthquake caused a particular damage pattern. The structuraldamage was confined to columns of the exterior frames. The rest ofthe observed damage was non-structural: cracking on the partitionwalls (inside the building) and the diagonal cracks of the masonryinfill walls at the exterior frames. The field observations indicatedthat the combined axial-shear failure at the top portions of the

columns were due commonly observed ‘‘short-column’’ effect thatresults from increased shear demands on a specific portion of thevertical members due to a decrease in the effective length. Several ofthe first-story columns of the Odontology building, which weredamaged due to short-column effect, are shown in Fig. 11. In thelight of the above described observations, it is deemed suitable tomodel the seven bay exterior frame of the building, as indicated inFig. 9.

4.1.3. Modeling approach

The fiber-based finite element analysis software ZEUS NL [5] isused to model the frame shown in Fig. 10. All members except for themasonry infill walls are represented with 3-D elasto-plastic beam-column elements described by a cubic shape function [9]. Thecolumns and slabs are modeled using rectangular sections. Theeffective flange width (for regular floors where a T-beam exists) is

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245 237

calculated according to (ACI 318-08 [1]). For the flat-slab floors, theportion of the slab that contributes to the frame analysis is deter-mined by using the formulations proposed by Luo and Durani [16,17].The concrete material is modeled using the constant confinementmodel developed by Martinez-Rueda and Elnashai [18]. Steel ismodeled with a bilinear elasto-plastic model with kinematic strainhardening. Based on the information on the design drawings, thefollowing material properties are used: a concrete compressivestrength of 29.4 MPa (300 kgf/cm2), steel yield and ultimatestrengths of 411.6 MPa (4197.2 kgf/cm2) and 617.5 MPa

7@6.3 m

3.10 m

1.87 m

3.85 m

1.65 m

3.85 m

1.25 m

Fig. 10. Elevation view sketch o

Fig. 11. Short-column effect observ

7@6.3 m = 44.1 m

5@7.

0 m

= 3

5 m

P6

P5P3 P2

P4 P1 P10

Wall B

Wall C

Wal

l A

Wal

l A

P6 P6

P6

P3

P3P3

P4

P4

P4

P2

P2

P2

P1 P1

P1

P5

P5

P5

P5 P5 P5 P5 P5 P5

P5 P5 P5 P5 P5 P5

P5

P5Wall B

Wall CWal

l A

Wal

l A

P10

Modeled FrameX

Y

N

Fig. 9. A typical plan view sketch of the Odontology building.

(6296.7 kgf/cm2), respectively. In modeling the masonry infill wallsthe following assumptions are made: infill walls do not carry anyvertical loads and they can be represented with diagonal struts thathave horizontal resistance only. The modeling of masonry infillwalls follow the approach proposed by Mostafaei and Kabeyasawa[19] as described in Kwon and Kim [13].

4.1.4. Analysis results

Two building configurations are considered: with and withoutinfill walls. An eigenvalue analysis indicates that the first andsecond mode periods of the building reduce from 0.83 s and 0.27 sto 0.80 s and 0.26 s respectively if infill walls are included in themodeling. It can be concluded that the shift in building periodwould not cause a significant change in the force demand on anequivalent single-degree-of-freedom structure; however, the pre-sence of the masonry infill walls increases the shear demand onthe first story columns as demonstrated in the following.

Inelastic dynamic time history analyses are conducted using atotal of eight ground motion records. As described previously inSection 2.3, for the sites of the case study structures three accelerationresponse spectra are developed using the source characteristics and aset of suitable attenuation relationships. For this case study, thespectrum compatible records are generated based on the originalrecordings CCSP and MELP. Additionally, for the back-analysis con-ducted here the original (without spectrum matching) record fromthe CCSP station (both horizontal components) is also used due toclose proximity of the station to the building site (6.7 km) asillustrated in Fig. 8(a).

The interstory drift ratio profiles for all eight records and thetwo configurations (with and without infill walls) are provided inFig. 12. It is observed that the interstory drift ratios are very

= 44.1 m

f the Odontology building.

ed at the Odontology building.

0 1 2 3

1

1.5

2

2.5

3

3.5

Interstory Drift Ratio (%)

Sto

ry

0 1 2 3Interstory Drift Ratio (%)

0 1 2 3Interstory Drift Ratio (%)

PSCC:drocerlanigirO PLEM:drocerelbitapmocmurtcepSSpectrum

compatible record: CCSP

Fig. 12. Maximum interstory drift ratio: dotted lines show frame without infill wall, solid lines show frame with infill wall.

10 20 30 40 500

100

200

300

400

500

Time (sec)

She

ar (k

N) Capacity

Demand

10 20 30 40 500

100

200

300

400

500

Time (sec)

She

ar (k

N)

Fig. 13. Comparison of shear capacity and demand for rightmost first story column, spectrum compatible mean record with seed CCSP: frame (a) without and (b) with

infill walls.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245238

similar for both cases except for the first story. The reduction inthe drift ratio at the first story (due to the presence of infill walls)ranges from 44 percent to 51 percent depending on the groundmotion.

The shear strength of the columns is calculated according to(ACI 318-08 [1]) and compared to the earthquake demandobtained from inelastic dynamic analysis. As an example, theshear capacity and the shear demand on the rightmost first storycolumn of the modeled frame is shown in Fig. 13. It is observedthat the presence of masonry infill walls significantly increase theshear demand on the columns and the shear capacity is exceededat certain times throughout the analysis. The maximum sheardemand to capacity ratio is illustrated in the bar plot in Fig. 14.It is concluded that the existence of masonry infill walls sig-nificantly increases the likelihood of exceeding the shear capacity.

It is concluded that the interstory drift ratio at the first story issignificantly reduced when the infill walls are present in theexpense of a significant increase in the shear demand. Manycolumns of the first story are expected to fail due to increasedshear demand as a result of the masonry infill walls. Particularly,for the exterior columns of the considered frame the sheardemand exceeds the shear capacity considerably which indicatessevere damage. The analytical results are in agreement with theobserved damage in the field.

4.2. Paso Cladio Arrau

The reference bridge (Paso Cladio Arrau) for this case study hasfour spans and it is 77.5 m long. The bridge has three bents eachof which consists of 10 piers as shown in Fig. 15. As observed inmost other bridges in Chile, seismic restrainers were installed inthe bridge. Girders are supported on elastomeric bearings made ofneoprene. The bridge has a skew angle of approximately 50degrees.

The bridge suffered minor damage from the earthquake asshown in Fig. 16. The superstructure displaced in the transversedirection, seismic restrainers yielded and elongated, and the shearkeys failed. Even though the shear keys were damaged, it is likelythat they prevented the unseating of the bridge superstructure. Inaddition to the shear keys, the seismic restrainers might havereduced the displacement demands as those can reduce the effectof vertical ground motion by maintaining the contact between thesuperstructure and neoprene bearings.

The 3D model of the bridge was built using ZEUS NL [5] asshown in Fig. 17(a). The stiffness of abutments and hystereticbehavior of the bearings are modeled using lumped springs. Thepiers, seismic restrainers, and superstructure are modeled usingcubic beam-column elements. Due to lack of information regard-ing the post-tensioning force in the seismic restrainers, it is

C1 C2 C3 C4 C5 C6 C7 C80

0.5

1

1.5

She

arD

eman

d/C

apac

ity

C1 C2 C3 C4 C5 C6 C7 C80

0.5

1

1.5S

hear

Dem

and/

Cap

acity

C1 C2 C3 C4 C5 C6 C7 C80

0.5

1

1.5

Columns

She

arD

eman

d/C

apac

ity

Fig. 14. Maximum shear demand to capacity ratio: empty bars indicate frame without infill walls, solid bars indicate frame with infill walls.

Fig. 15. Configuration of bridge bents and abutment.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245 239

assumed that they were post-tensioned up to 30 percent of theyield stress. The additional normal force to elastomeric bearingsdue to the post-tensioning is taken into account in the hystereticbehavior of elastomeric bearings; the hysteretic curve is shown inFig. 17(b). The deck is considered as a rectangular beam havingthe equivalent section properties of the girders and the bridgedeck. The configuration of the superstructure–substructureconnections is illustrated in Fig. 17(b).

Spectrum compatible ground motions from Section 2.3 areused for inelastic response history analyses. A total of 18horizontal components of ground motions are applied in thetransverse direction. Two sets of analyses are carried out. In thefirst set, the seismic restrainers are included in the numericalmodel as flexural elements. In addition, the normal force on theelastomeric bearings is increased to take into account the post-tensioning force. It is assumed that the additional normal force onelastomeric bearings due to post-tensioning remains constantthroughout the earthquake excitation. For the second set ofanalyses, all parameters are kept the same as those in the firstset except that the contribution of seismic restrainers is removedfrom the model.

To illustrate the modeling of gaps at the both ends of thebridge, the trajectories of the bridge deck end points are shown inFig. 18. If the bridge span moves only in the transverse directionwithout rotation, Node A in Fig. 17(a) is bounded between twosolid lines in Fig. 18, which indicate the maximum displacementlimits corresponding to the cases where the gaps at the ends areclosed. If there is a rotational component, Node A may movefurther away from the upper boundary. Since the sample resultshown in the figure is from a low-intensity input motion, whichdoes not develop large rotational displacement, the trajectory ofNode A stays within the boundary. Fig. 19 compares the max-imum transverse displacement of Node A for the two analysiscases as a function of PGA of the input ground motion. As it can beobserved from the figure, the models without seismic restrainersgenerally have larger displacement demand, and the trend isclearer as the PGA of the input ground motions increases. Thisdifference is primarily due to the increased normal force inelastomeric bearings, which increases the energy dissipationcapacity of the bearings.

4.3. Las Mercedes Bridge

4.3.1. Configuration, observed damage and numerical model

Las Mercedes Bridge exemplifies the most commonly usedhighway overpass design in Chile in the recent years. Severalbridges with similar configuration suffered damage during theearthquake. The damage to the bridge, mainly due to excessivemovement and rotation of the superstructure in the transversedirection resulting in the failure of the shear keys and unseatingof the deck, is shown in Fig. 20. The bridge has a two-spancontinuous deck with a span length of 28.5 m. The superstructureconsists of pre-stressed girders and cast-in-place reinforced con-crete slabs and is supported by a two-column bent at the centerand retaining walls at both ends as shown in Fig. 21(a). Thelongitudinal direction of the bridge is skewed by an angle of 11degrees from the perpendicular direction to the highway which itpasses over.

Fig. 17. Details of the numerical model.

Fig. 16. Observed damage to Paso Cladio Arrau.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245240

Three-dimensional numerical models are generated by usingthe finite element analysis program ZEUS NL [5] to investigate theseismic response of the bridge. Based on the design specificationof a typical bridge on the highway (Route 5) and Schmidt hammertests at the field, compressive strength of concrete is assumed tobe 50 MPa (510 kgf/cm2) for girders and 35 MPa (357 kgf/cm2) forother parts of the bridge including the columns. The yield stress of

reinforcement is assumed as 400 MPa (4079 kgf/cm2). Overviewof the numerical model is provided in Fig. 21(b).

Elastomeric bearings are placed at both ends of each girderand a total of twelve bearings are used as shown in Fig. 22(a). Theforce–displacement relationship of the bearing is determined byfriction force and friction displacement; see Fig. 22(b). The frictioncoefficient between the surfaces of the bearing and the girder is

Fig. 18. Trajectory at the end of bridge span.

Fig. 19. Maximum responses of the bridge with (w/) and without (w/o) seismic

restrainers.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245 241

assumed to be 0.21 which is the friction coefficient betweenneoprene rubber and concrete. The thickness of bearing pads onthe column bent is 37 mm, while the thickness of those onretaining walls is 60 mm. The maximum friction displacement(Db) at which the friction slip begins is assumed to be 40 mm and25 mm for the tall and short bearings, respectively. Gap elementsare used to represent the delayed contact at the constructionjoints between the bridge deck and approach slabs on theabutments. Based on the design drawings, the clearance of theconstruction joint is assumed to be 60 mm. A unidirectionalspring element is used to represent the conditional force transferat the construction joint; only compressive forces are transferredwhen the gap is closed.

4.3.2. Analysis results

Under earthquake excitations, the bridge behaves as a semi-rigid diaphragm on a stiff substructure. The response of thesuperstructure is mainly determined by slip at the girder-bearinginterface. The fundamental period of the bridge is evaluated as0.80 s and this mode is induced by the deformation of theelastomeric bearing before the friction slip. The first mode shapeis determined by the movement of the superstructure in thetransverse direction without torsional responses because thebridge is symmetric along its length. In the static pushoveranalysis, lateral forces are applied in the transverse direction.The locations of shear keys and direction of lateral forces areshown in Fig. 23(a). From the analysis results, it was observedthat shear key #1 reaches its ultimate shear capacity earlier thanothers, followed by severe damage on shear key #2. The reason isthat the shear keys on the stiff embankments (#1 and #3) attract

more forces than shear key #2, which is placed on the flexiblecolumn bent. Under the lateral forces that are assumed for thestatic pushover analysis [see Fig. 23(a)], shear key #1 reaches itsultimate capacity earlier than shear key #3 because the center oflateral forces shifts to the left. This shift is caused by a minordistortion of the superstructure due to large lateral forces. Unevenlateral force distribution on shear keys and unbalanced forceredistribution can magnify torsion in highly inelastic responserange of the bridge. The status of the bridge on the capacity curvein Fig. 23(b) is explained as follows: (1) the superstructuredisplaces due to lateral deformation of elastomeric bearings;(2) the lower part of a girder hits a shear key; (3) shear key #1reaches its ultimate shear strength; (4) force redistribution occursamong the shear keys; (5) shear key #2 reaches its ultimate shearstrength. As soon as the shear key fails, the girder and elastomericbearing slip off from the bearing support and once the latter slip-off occurs, the girder hardly moves back in the opposite directionand thus its displacement accumulates unidirectionally. Atthe same time, the other end of the superstructure moves in theopposite direction, which causes cumulative torsion of thesuperstructure.

A sample plot of dynamic response history analysis by therecord matched to þ0.5s spectrum using the original groundmotion from the station STL (EW component), see Section 2.3, isshown in Fig. 24. The earthquake loading is applied in thetransverse direction of the bridge deck. Shear keys are omittedin the analytical model in order to investigate causes of torsionother than the uneven resistance of shear keys. While themaximum displacement is 90 mm, which is larger than theallowable deformation of the elastomeric bearing, the torsion isnegligible, only about 0.0012 degrees. This implies that the largeamount of torsion observed at the field had not been caused bydynamic characteristics of the structure. In this reference bridge,the skew angle is too small for the observed torsion to be causedby the pounding response. Based on the above discussion, thelarge torsion observed at the field is explained as follows:premature failure of a shear key causes uneven and excessivedisplacement at an end of the superstructure. Then the girder endslips off from the bearing support and this initiates the torsionthat is accumulated unidirectionally.

4.3.3. Fragility curve development

A multi-component fragility analysis is performed on this casestudy bridge using an analytical fragility methodology, whichincludes using joint probabilistic seismic demand models tointegrate the responses of several bridge components in the

Fig. 21. (a) Configuration of the Las Mercedes bridge (plan view) and (b) overview of the numerical model.

Fig. 22. Numerical model: (a) locations of joint and gap elements and (b) properties of joint elements for bearings.

Fig. 20. Damage to Las Mercedes Bridge.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245242

analysis. The monitored components include the column ducti-lity, column maximum drift ratio, and abutment movement in thelongitudinal and transverse directions. The limit states used forthe fragility curve generation are adopted from the research ofNielson and DesRoches [21], and Ramanathan et al. [23]. Toaddress the issue of uncertainty for this analysis, the bridgegeometry and material properties are modeled deterministically,and only the suite of ground motions chosen contributed to thevariability in the response. This is because the variability of theresponse of the system is much more susceptible to the groundmotion variability than the material uncertainties [12]. The suiteof ground motions chosen for this analysis characterizes the

specified region and also provides a range of intensities. Theprocurement and development of these ground motions isdescribed in Section 2.3.

Fig. 25 shows the fragility curves for this bridge based on thepeak ground acceleration (PGA). Two fragility curves are shownfor the same bridge, one curve using column curvature ductilityas the engineering demand parameter for columns, and the otherusing maximum column drift ratio, Fig. 25(a) and (b) respec-tively. As is shown, these two parameters give slightly differentestimates of the fragility of this bridge in all of the damagestates. For example, for a PGA level of 0.5 g, the fragility curvebased on curvature ductility predicts around 50 percent chance

0 20 40 60 80 100-100

-50

0

50

100

Time (sec)

Dis

plac

emen

t (m

m)

0 20 40 60 80 100-0.002

-0.001

0

0.001

0.002

Time (sec)

Tors

iona

l ang

le (d

egre

e)

Fig. 24. Selected dynamic response history results: (a) transverse displacement at the center of the bridge and (b) angle of torsion.

Shear key #1

Shear key #2

Shear key #3

Equivalent lateral force

Equivalent lateral force

0 50 100 150 200 250 3000

1000

2000

3000

4000

5000

Displacement (mm)

Bas

e sh

ear (

kN)

(1) (2)

(3)(4)

(5)

Fig. 23. (a) Direction of equivalent lateral forces and locations of shear keys and (b) capacity curve of the reference bridge.

0 0.5 1 1.5 20

0.2

0.4

0.6

0.8

1

PGA (g)

P(L

S|P

GA

) SlightModerateExtensiveComplete

0 0.5 1 1.5 20

0.2

0.4

0.6

0.8

1

PGA (g)

P(L

S|P

GA

)

Fig. 25. Fragility curves based on (a) column curvature ductility, and (b) maximum column drift ratio.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245 243

of slight damage, while the drift ratio based curve predictsapproximately 20 percent probability of slight damage. Thisdifference could be due to the fact that the limit states comefrom different sources demonstrating that the limit states usedin fragility analysis affect the results greatly. These curves showthat the bridge has low vulnerability to damage for the moresevere limit states.

The damage prediction by the fragility curves provided inFig. 25 are in accordance with the field observations in thatalmost no damage was observed in the bridge columns due tolimited continuity with the superstructure. The failure resultingfrom the demolition of the shear keys and the unseating of thedeck is addressed in the back analysis described in the previoussections.

A.S. Elnashai et al. / Soil Dynamics and Earthquake Engineering 42 (2012) 229–245244

5. Summary and conclusions

An earthquake of magnitude 8.8 hit central Chile on February27, 2010. Despite the large magnitude and proximity of the eventto large settlements, the majority of engineered buildings per-formed satisfactorily and the loss of life was limited. However, thetransportation system, hospitals and schools were affected sig-nificantly causing disruption in the aftermath of the earthquake.The economic losses were major amounting to approximately 17percent of the GDP of Chile.

Large magnitude earthquakes that affected densely populatedregions with well-designed engineered structures are sparse.Therefore, the 2010 Maule (Chile) earthquake is an importantopportunity for the scientific community to study the strongground motion produced by such events and the deficiencies indesign practices to prevent the impacts of future earthquakes. Inthis paper, due to the lack of strong ground motion data, hazardand site specific acceleration time histories are developed throughspectrum matching of response spectra obtained from selectedattenuation relationships. The developed ground motions areused for back-analysis of a building and two bridges to investigatethe dynamic response of these structures under the large magni-tude earthquake event and to explain the commonly observedmodes of failure. For the building, the analysis results confirmedthe shear failure of column observed in the field resulting fromthe increased demand due to presence of infill walls. This type offailure is also known as the short-column effect and it is widelyseen in the past earthquakes as well. The bridge case studiesaimed at understanding the effectiveness of seismic restrainers(that are peculiar to Chilean bridge construction) and unseating ofbridge decks due to shear key failures extensively observedduring the 2010 Maule earthquake. The analysis results indicatethat the seismic restrainers are an effective way of reducing thedisplacements of bridge decks and their effectiveness increaseswith increasing ground motion intensity. It is also found out thatthe torsional response of bridge superstructure is a result ofunidirectional accumulation of displacements that occur after thefailure of weak shear keys. It is concluded that if shear keys hadbeen designed properly to resist the loads from superstructure,damage to bridges could be minimized and several of the bridgeswould be functional after the earthquake, facilitating the rescueand response efforts.

Acknowledgments

The field mission to Chile was sponsored by the Mid-AmericaEarthquake Center, Missouri University of Science and Technology,University of Connecticut, Georgia Institute of Technology, andNational Research Foundation Grant provided by the Governmentof Korea (2011-0028552). The MAE Center is a graduated NationalScience Foundation (NSF) Engineering Research Center, which wasfunded under NSF Grant EEC-9701785. The authors express theirgratitude to the following individuals: Jeffery Roesler, Imad L.Al-Qadi, Angharad Valdivia, Rafael Riddell, Guillermo Thenoux Z.,Marcelo Gonzalez H., Carlos Videla, Mauricio Lopez, MauricioPradena Miquel, Ramon Verdugo, Gregory Pluta, Carolina Cerda,Juan Vargas, Luis Echeverria, Moises Vargas Eyzaguirre, FernandoGonzalez, and Jonguen Baek.

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